18th International Conference On Soil Mechanics, Paris 2013 PDF [PDF]

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Contents / Table des matières Foreword Avant-propos

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ISSMGE – The State of the Society (2009-2013) SIMSG – État de la Société (2009-2013) Briaud J.-L.

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Honour Lectures / Conférences honorifiques 8th Terzaghi Oration - Protecting society from landslides – the role of the geotechnical engineer 8e allocution Terzaghi - La gestion de l’aléa glissements de terrain et le rôle de l’ingénieur géotechnicien Lacasse S.

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Bishop Lecture - Advanced laboratory testing in research and practice Conférence Bishop - Les essais en laboratoire avancés dans la recherche et dans l’industrie Jardine R. J.

35

Ishihara Lecture - Soil–Foundation–Structure Systems Beyond Conventional Seismic Failure Thresholds Conférence Ishihara - Les systèmes sol-fondation-structure qui dépassent les limites de la rupture parasismique conventionnelle Gazetas G.

55

Kerisel Lecture - The role of Geotechnical Engineers in saving monuments and historic sites Conférence Kerisel - Le rôle des ingénieurs géotechniciens dans la sauvegarde des monuments et des sites historiques Calabresi G.

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McClelland Lecture - Analytical contributions to offshore geotechnical engineering Conférence McClelland - Contributions des méthodes analytiques à la géotechnique offshore Randolph M. F.

85

Ménard Lecture - The pressuremeter test: Expanding its use Conférence Ménard - L’essai pressiometrique : élargissement de son utilisation Briaud J.-L.

107

Rowe Lecture - The role of diffusion in environmental geotechnics Conférence Rowe - Le rôle de la diffusion en géotechnique environnementale Shackelford C.

127

Schofield Lecture - Centrifuge modelling: expecting the unexpected Conférence Schofield - Modélisation physique en centrifugeuse : prévoir l’imprévisible Bolton M. D.

151

Special Lectures / Conférences spéciales Enjeux géotechniques pour la construction du métro automatique « Grand Paris Express » Geotechnical issues for « Grand Paris Express » automatic metro Fluteaux V.

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Innovations françaises en géotechnique: les projets nationaux de recherche French Innovations in Geotechnics: the National Research Projects Schlosser F., Plumelle C., Frank R., Puech A., Gonin H., Rocher-Lacoste F., Simon B., Bernardini C.

163

The new Bugis Station and associated tunnels for the Singapore MRT Métro de Singapour : nouvelle station Bugis et tunnels associés Sim A.

183

Technical Committee 101 Laboratory Stress Strain Strength Testing of Geomaterials Session I - Time effects and other peculiar observations Comité technique 101 Caractérisation en laboratoire du comportement des géomatériaux Session I - Effets du temps et autres aspects General Report of TC 101 - Session I - Laboratory testing of geomaterials: Time effects and other peculiar observations Rapport général du TC 101 - Session I - Essais de laboratoire sur les géomatériaux : effets du temps et autres observations spécifiques Ibraim E. Engineering properties of an expansive soil Propriétés mécaniques d’un sol gonflant Azam S., Ito M., Chowdhury R.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Methods of determination of K0 in overconsolidated clay Méthodes de détermination de K0 dans une argile surconsolidée Boháč J., Mašín D., Malát R., Novák V., Rott J.

203

Stress-path effects on the grading of an artificial material with crushable grains Stress-trajectoire effets sur le granulométrie d’un matériau artificiel avec des grains déformables Casini F., Leu J., Low B., Wanninger F., Zimmermann A., Zwicker P., Springman S.M., Viggiani G.M.B.

207

Characteristics of structure evolution of expansive soil and loess during loading and wetting Caractéristiques de l’évolution structurale du sol expansif et du loess lors du chargement et du mouillage Chen Z.

211

Misconceptions about experimental substantiation of creep hypothesis A Les idées fausses justifiant l’hypothèse A de fluage au laboratoire Degago S.A., Grimstad G., Jostad H.P., Nordal S.

215

The relationship between swelling and shear strength properties of bentonites La relation entre les propriétés de résistance au cisaillement de l’enflure et des bentonitiques Domitrović D., Kovačević Zelić B.

219

Reappraisal of Surcharging to Reduce Secondary Compression Remise en cause de l’imposition de frais supplémentaires pour réduire la Compression secondaire Feng T.W.

223

Effets d’échelle dus à la rupture des grains sur la résistance au cisaillement d’enrochements Size effects due to grain crushing in rockfill shear strength Frossard E., Ovalle C., Dano C., Hicher P-Y., Maiolino S., Hu W.

227

Electro-osmotic consolidation: Laboratory tests and numerical simulation Électro-osmotique de consolidation : les tests de laboratoire et simulation numérique Hu L., Wu H., Wen Q.

231

Creep characteristics of clay in one-dimensional compression with unloading/reloading cycles Propriétés de fluage des argiles en compression unidimensionnelle avec cycles de charge/décharge Kawabe S., Tatsuoka F.

235

Comparison of the geotechnical properties of pumice sand from Japan and New Zealand Comparaison des propriétés géotechniques de sables de pierre ponce du Japon et de Nouvelle-Zélande Kikkawa N., Pender M.J., Orense R.P.

239

Evaluation of Consolidation Behavior of Soils under Radial Drainage Condition Using Digital Image Analysis Évaluation du comportement de consolidation des sols sous des conditions de drainage radial à partir de l’analyse d’image numérique Kim J.-Y., Chung C.-K., Cho N.-G., Yune C.-Y.

243

Mise au point d’un dispositif expérimental pour l’analyse du retrait-gonflement des sols argileux Development of an experimental device for swelling-shrinkage analysis of clayey soils Maison T., Laouafa F., Delalain P.

247

Residual shear strength behavior of swelling soils Comportement de force résiduelle de cisaillement des sols gonflants Markou I.N.

251

Rational expression of time-dependent behavior from normally consolidated soil to naturally deposited soil Expression rationnelle du comportement dépendant du temps des sols normalement consolidés et déposés naturellement Nakai T., Shahin H.M., Kyokawa H.

255

Quantification des gonflements des sols surconsolidés en fond de déblai Swelling quantification of overconsolidated soils at excavation base Petit G., Virollet M., Bernard Ph.

259

Rate effects at varying strain levels in fine grained soils Effets de vitesse de déformation à niveaux de déformation variant en sols à grains fins Robinson S., Brown M.J.

263

Comparison of Stress-Strain Behaviour of Carbonate and Silicate Sediments Comparaison de la réponse contrainte-déformation de sédiments carbonatés et siliceux Safinus S., Hossain M.S., Randolph M.F.

267

A new test field in sulphide clay with test embankments for study of compression properties Un nouveau essai sur le terrain d´argile sulfaté en mettant en place des remblais d´essai pour l’étude des propriétés de compression Westerberg B., Andersson M.

271

Laboratory testing issues related to crushable sands Questions concernant des essais de laboratoire sur les sables écrasables Wils L., Van Impe W.F., Haegeman W., Van Impe P.O.

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Contents / Table des matières

Non-coaxial behaviour of sand in drained rotational shear Comportement non-coaxial de sable drainé en cisaillement rotationnel Yang L.-T., Yu H.-S., Wanatowski D., Li X.

279

Test study and constitutive modelling of the time-dependent stress-strain behavior of soils Test et modélisation du comportementen fonction du temps de contrainte-déformation comportement des sols Yin J.-H., Tong F.

283

Special Features of Creep of Clayey Soils Particularités du fluage des sols argileux Zhakulin A.S., Zhakulina A.A., Orazaly E.E., Orazalin Z.Y.

287

Technical Committee 101 Laboratory Stress Strain Strength Testing of Geomaterials Session II - Strength properties and treated soils Comité technique 101 Caractérisation en laboratoire du comportement des géomatériaux Session II - Propriétés de résistance et sols traités General Report - Session II - Laboratory Testing of Geomaterials: Strength Properties and Treated Soil Rapport général - Session II - Essais de laboratoire des géomatériaux : propriétés mécaniques et sols traités Kim D.-S.

293

Triaxial testing of asphalt Essais triaxiaux de l’asphalte Airey D., Prathapa R.

301

Bounding surface plasticity model parameters for Bagdad soils Paramètres du modèle de plasticité de surface de délimitation pour les sols de Bagdad Al-Farouk O., Al-Damluji S., Al-Shakarchi Y.J., Albusoda B.S.

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The December 29th 2010 Xerolakka Municipal Solid Waste landfill failure 29 décembre 2010 : l’échec d’enfouissement Xerolakka Athanasopoulos G., Vlachakis V., Zekkos D., Spiliotopoulos G.

309

Shear Strength and Deformation Modulus of Tailing Sands under High Pressures Résistance au cisaillement et module de déformation de sables de rejets sous hautes pressions. Campaña J., Bard E., Verdugo R.

313

A Comparison Between the Shear Strength Measured with Direct Shear and Triaxial Devices on Undisturbed and Remolded Soils Une comparaison entre la résistance au cisaillement mesurée avec appareils de cisaillement direct et triaxiaux sur les sols non remaniés et remoulés Castellanos B.A., Brandon T.L.

317

Experimental analysis of the mechanical properties of artificially cemented soils and their evolution in time Analyse expérimentale des propriétés mécaniques des sols cimentés artificiellement et leur évolution dans les temps Consoli N.C., Fonini A., Maghous S., Schnaid F., Viana da Fonseca A.

321

Influence of diatom microfossils on soil compressibility Influence des microfossiles de diatomées sur la compressibilité des sols Díaz-Rodríguez J.A., González-Rodríguez R.

325

Strength properties of densely compacted cement-mixed gravelly soil Ppropriétés de résistance des graves cimentées fortement compactées Ezaoui A., Tatsuoka F., Furusawa S., Hirao K., Kataoka T.

329

Tensile Strength of Lightly Cemented Sand through Indentation Tests Résistance à la traction de sable légèrement cimenté par des tests d’indentation Ge L., Yang K.-H.

333

Mechanisms During Formation of Ice Lenses and Suction in Freezing Soils Les mécanismes de la formation des lentilles de glace et de succion au cours de la congélation du sol Herzog F., Boley C.

337

Comportement en petites déformations d’un sol traité à la chaux Small strain behavior of a lime-treated soil Hibouche A., Taibi S., Fleureau J.-M., Herrier G.

341

A Key Parameter for Strength Control of Lightweight Cemented Clays Un paramètre clé pour le contrôle des forces de légères argiles cimentées Horpibulsuk S., Suddeepong A., Chinkulkijniwat A.

345

Some notes concerning the dry density testing standards Quelques remarques concernant les descriptions relatives aux essais de densité sèche Imre E., Lőrincz J., Gerendai E, Szalkai R, Lins Y., Schanz T.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Interpretation of stress-dependent mechanical behaviour of rockfill materials Interprétation de stress-dépendante et comportement mécanique de matériaux enrochement Jannati Aghdam R., Soroush A.

353

Effects of Freeze-Thaw History on Deformation-Strength Properties and Permeability of Fine-Grained Soil Effets de l’historique du gel-dégel sur les propriétés de résistance à la déformation et de perméabilité des sols à grains fins Kawaguchi T., Nakamura D., Yamashita S., Yamasaki S., Ishikawa T.

357

Characterization of geomechanical and hydraulic properties of non-wettable sands Caractérisation des propriétés géomécaniques et hydrauliques des sables non mouillants Kim D., Yang H.-J., Yun T.S., Kim B., Kato S., Park S.-W.

361

The strength change characteristics of weathering soil due to repeat freezing-thawing and drying-wetting Les caractéristiques de changement de résistance du sol aux intempéries suite aux répétitions de la congélation-décongélation et du séchage-amortissement Kim Y.S., Seong J.H., Kim S.S

365

Soil structure in gravel-mixed sand specimen and its influence on mechanical behavior Structure du sol des échantillons de sable avec gravier et son influence sur le comportement mécanique Kodaka T., Cui Y., Mori S., Kanematsu Y., Lee K.-T.

369

The expansive properties of Poland’s clay subsoil Propriétés de l’argile expansive de substrat de la Pologne Kumor M. K.

373

Effect of Particle on K0 Behaviour for Granular Materials Effet des caractéristiques particulaires sur le comportement des matériaux granulaires K0 Lee J., Park D., Kyung D., Lee D.

377

Duncan-Chang - Parameters for Hyperbolic Stress Strain Behaviour of Soft Bangkok Clay Duncan-Chang - Paramètres de comportement contrainte-déformation hyperbolique d’argile molle de Bangkok Likitlersuang S., Surarak C., Balasubramania A., Oh E., Syeung Ryull K,. Wanatowski D.

381

Laboratory investigation of seismic effects of nanoparticle dispersions in saturated granular media Étude en laboratoire des effets sismiques des dispersions de nanoparticules dans les milieux granulaires Luke B., Werkema D., Andersen S.

385

The SCS Double Hydrometer Test in dispersive soil identification Essai SCS de double hydrométrie pour l’identification des sols dispersifs Maharaj A., Paige-Green P.

389

Correlation between deflections measurements on flexible pavements obtained under static and dynamic load techniques Corrélation entre les déflexions de revêtements flexibles mesurées sous chargement statique et dynamique Murillo Feo C.A., Bejarano Urrego L.E. Comparison of permeability testing methods Comparaison des différentes méthodes sur les tests de perméabilité Nagy L., Takács A., Huszák T., Mahler A., Varga G. Oscillation of Acceleration Accompanying Shear Band and Subsequent Time-Dependent Behavior in Overconsolidated Clay under Undrained Plane-Strain Conditions Oscillation de l’accélération accompagnant la formation de bandes de cisaillement et comportement dépendant du temps dans une argile surconsolidée en déformations planes et conditions non drainées. Noda T., Xu B.

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Behavior of fine-grained soils compacted with high shear stresses Comportement des sols fins compactés avec des niveaux de cisaillement élevés Perez N., Garnica P., Mendoza I., Reyes M.A

407

Influence of Minerals on the Elastic Behaviour of Cohesive Soil Influence des minéraux sur le comportement élastique des sols cohésifs Sarma D., Sarma M.D.

411

Experimental Analysis on the Influence of Surcharge Filters on Safety Against Hydraulic Heave Analyse expérimentale de l’influence d’un filtre de surcharge sur la stabilité contre des soulèvements d’eau d’une fouille de construction Schober P., Boley C.

415

Coupled THM mechanical model for porous materials under freezing condition Couplé THM modèle mécanique pour les matériaux poreux dans des conditions de congélation Shin H., Ahn J.-H., Kim Y.-T., Lee S.-R.

415

Correlation between drained shear strength and plasticity index of undisturbed overconsolidated clays Corrélation entre la résistance au cisaillement des sols drainés et l’indice de plasticité des argiles surconsolidés non perturbées Sorensen K.K., Okkels N.

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Contents / Table des matières

Mechanisms of binder interactions and their role in strengthening Kuttanad clay Mécanismes d’interaction des liants et leur rôle dans le renforcement de l’argile de Kuttanad Suganya K., Sivapullaiah P.V.

429

Hardening process of clayey soils with high water content due to thixotropy effect Processus de durcissement des sols argileux à forte teneur en eau causé par un effet thixotropique Tanaka H., Seng S.

433

Comparative study of long-term consolidation for subsoils under Kansai Airport and Pisa Tower Étude comparative de la consolidation à long terme pour les sous-sols d’aéroport de Kansai et de tour de Pise Watabe Y., Sassa S., Udaka K.

437

Settlement and shear strength of uncemented coal mine overburden materials placed loose under dry and wet conditions Tassement et résistance au cisaillement de matériaux de couverture non cimentés extraits de mines de charbon et mis en dépôt en vrac dans des conditions sèches et humides Williams D.J., Kho A.K.

441

Anisotropic strength ratio and plasticity index of natural clays Étude de la relation entre l’anisotropie et l’indice de plasticité des argiles naturelles Won J.Y.

445

Hydraulic Heave in Cohesive Soils Rupture hydraulique du sol en terrain cohérent Wudtke R.-B., Witt K.J.

449

Evaluation of sample disturbance due to the exsolution of dissolved gas in the pore water of deep lake bottom sediments Évaluation du remaniement des échantillons dû à l’exsolution de gaz dissous dans les eaux interstitielles des sédiments de fond de lacs profonds Yamashita S., Miura R., Kataoka S.

453

Fabric and critical state of granular materials La structure et l’état critique des matériaux granulaires Yan W.M., Zhang L.

457

Study on New Method of Accelerated Clay Creep Characteristics Test Étude d’une nouvelle méthode d’évaluation accélérée des caractéristiques de fluage des argiles Ye Y., Zhang Q., Cai D., Chen F., Yao J., Wang L.

461

Constitutive model and simulation of non-segregation freezing and thawing in soils Modèle de comportement et simulation du gel et le dégel des sols sans ségrégation Zhang Y., Michalowski R.L.

465

Technical Committee 102 Ground Property Characterization from In-Situ Tests Comité technique 102 Caractérisation des propriétés des terrains par essais in situ General Report for TC102 In-Situ Testing Rapport général du TC102 Essais in-situ Giacheti H.L., Cunha R.P.

471

Challenging Problems of Gypseous Soils in Iraq Des problèmes difficiles des sols gypseux en Irak Al-Saoudi N.K.S., Al-Khafaji A.N., Al-Mosawi M.J.

479

Site characterization by seismic dilatometer (SDMT): the Justice Court of Chieti Caractérisation du site par dilatomètre sismique (SDMT) : la cour de Justice de Chieti Amoroso S., Totani F., Totani G.

483

Détermination du coefficient rhéologique a de Ménard dans le diagramme Pressiorama®. 487 Obtaining the Ménard a Rheological Factor in a Pressiorama® Diagram Baud J.-P., Gambin M. Courbes hyperboliques contrainte-déformation au pressiomètre Ménard autoforé Stress-Strain Hyperbolic Curves Obtained With a Selfboring Ménard PMT Baud J.-P., Gambin M., Schlosser F.

491

Quality control of Cutter Soil Mixing (CSM) technology – a case study Contrôle de la qualité des la technologie Cutter Soil Mixing (CSM) – une étude de cas Bellato D., Simonini P., Grisolia M., Leder E., Marzano I.P.

495

Mesures dynamiques lors du battage pénétromètrique – Détermination de la courbe charge-enfoncement dynamique en pointe Dynamic measurements of the penetration test – Determination of the tip’s dynamic load-penetration curve Benz M.A., Escobar E., Gourvès R., Haddani Y., Breul P., Bacconnet C.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Performance of a deep excavation in downtown Toronto Performance d’une excavation profonde au centre-ville de Toronto Cao L.F., Peaker S.M., Ahmad S.

503

Permeability scale effect in sandy aquifers: a few case studies Effet d’échelle et perméabilité des aquifères sableux : quelques études de cas Chapuis R.P.

507

A Study of Cuttability Indices for Tunnel Penetration Étude sur les indices d’aptitude à la coupe pour la pénétration de tunnels Chen L.-H., Chen Y.-C., Chen W.-C., Liu H.-W.

511

Survey results of damaged areas in flood disaster of typhoon Morakot and suggestions for restoration projects Résultats des investigations sur les zones ravagées et inondées par le typhon Morakot, propositions de projets de restauration Chou J.C., Huang C.R., Shou K.J.

515

Stability of chilean’s tailings dams with the Panda® penetrometer. Experiences of the last 10th Dix ans d’études de la stabilité des barrages de résidus miniers chiliens à l’aide du pénétromètre Panda® Espinace R., Villavicencio G., Palma J., Breul P., Bacconnet C., Benz M.A., Gourvès R.

519

Site Sampling: Assessing Residual Uncertainty Échantillonnage du site : évaluation de l’incertitude résiduelle Fenton G.A., Hicks M.A.

523

Multi-Sleeve Axial-Torsional-Piezo Friction Penetration System for Subsurface Characterization Système de pénétromètre à friction axial-torsional-piezométrique à manchons multiples pour la reconnaissance des sols superficiels Frost J. D., Martinez A.

527

Hydraulic Properties of Glacial Deposits Based on Large Scale Site Investigation Les propriétés hydrauliques des dépôts glaciaires basées sur une enquête de chantier à grande échelle Galaa A., Manzari M., Hamilton B.

531

The seismic SPT test in a tropical soil and the G0/N ratio L’essai SPT sismique pour le sol tropicaux et la relation G0/N Giacheti H.L., Pedrini R.A.A., B.P. Rocha B.P.

535

Compressibility Parameters of Cohesive Soils From Piezocone Paramètres de compressibilité de sols cohésifs au piézocône Hamza M., Shahien M.

539

Comportement de la structure de sol amélioré par inclusions rigides, supportant une éolienne Behaviour of soil foundation improved by rigid columns, supporting a wind turbine Haza-Rozier E., Vinceslas G., Le Kouby A., Crochemore O.

543

Seismic Response of Superstructure on Soft Soil Considering Soil-Pile-Structure Interaction Influence de l’Interaction sol- pieu- structure sur la réponse sismique de la superstructure sur sol mou Hokmabadi A.S., Fatahi B., Samali B.

547

Applicability of the RNK-method for geotechnical 3D-modelling in soft rocks Applicabilité de la RNK-méthode pour la modélisation géotechnique en 3D en roches tendres Ivšić T., Ortolan Ž., Kavur B.

551

Une nouvelle sonde permettant de mesurer sans extrapoler la pression limite pressiométrique des sols A new probe for measuring the pressuremeter limit pressure of soils without extrapolation Jacquard C., Rispal M., Puech A., Geisler J., Durand F., Cour F., Burlon S., Reiffsteck P.

555

Long-term Deformation of the Reclaimed Pleistocene Foundation of the Offshore Twin Airport Déformations à long terme d’une fondation de remblai pléistocène récupéré sur mer pour un projet d’aéroport jumelé Jeon B.G., Mimura M.

559

Assessment of Scour Potential of a Circular Pier in Silty Sand Using ISEEP Caractérisation par ISEEP du potentiel d’érosion d’une pile circulaire dans un sable silteux Kayser M., Gabr M.

563

Practical Reviews on CO2 Sequestration in Korean Sedimentary Basins and Geophysical Responses of CO2-injected Sediments Le comportement pratiques sur la séquestration du CO2 dans les bassins sédimentaires coréens et réponses géophysiques de CO2 injectées sédiments Kim A.R., Cho G.C., Kwon T.H., Chang I.H.

567

Using Multi-scale Sediment Monitoring Techniques to Evaluate Remediation Effectiveness of the Tsengwen Reservoir Watershed after Sediment Disasters Induced by Typhoon Morakot 571 Utilisation des techniques de surveillance des sédiments mulit-échelles pour évaluer l’efficacité d’assainissement du bassin hydrographique du réservoir Tsegwen après les catastrophes de sédiments induites par le typhon Morakot Lin B.-S., Ho H.-C., Hsiao C.-Y., Keck J., Chen C.-Y., Chi S.-Y., Chien Y.-D., Tsai M.-F.

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Contents / Table des matières

Practice and development of the piezocone penetration test (CPTu) in geotechnical engineering of China La pratique et le développement de pénétration au piézocône (CPTu) en Chine Liu S., Cai G., Du Y., Puppala A.J.

575

The use of hydro test results for design of steel tanks on stone column improved ground - a case history L’emploi des résultats des essais hydrauliques dans l’étude des réservoirs en acier sur le sol amélioré par colonnes de pierre – histoire de cas Matešić L., Mihaljević I., Grget G., Kvasnička P.

579

Interrelationship between deformation moduli from CPTU and SDMT tests for overconsolidated soils La corrélation entre le module de déformation de CPTU et de tests SDMT pour les sols surconsolidés Młynarek Z., Gogolik S., Sanglerat G.

583

Le Géomécamètre, un nouvel essai in situ adapté à la mesure des caractéristiques hydro-mécaniques du sol The Geomechameter test, a new in-situ apparatus adapted to the measurement of the hydro-mechanical characteristics of the soil Monnet J.

587

Analytical approach for determining soil shear strength parameters from CPT and CPTu data Approche analytique pour déterminer le cisailler du sol et son paramètre de la résistance de CPT et CPTu data Motaghedi H., Eslami A., Shakeran M.

591

Use of penetration testing for determination of soil properties in earth dam Emploi des essais de pénétration pour déterminer les propriétés de sol pour barrages en terre Mulabdic M.

595

Diagnosis of earth-fills and reliability-based design Diagnostic de remblais de terre et conception basée sur la fiabilité Nishimura S., Shuku T., Suzuki M.

599

Correlation between cone penetration rate and measured cone penetration parameters in silty soils Corrélation entre le taux de pénétration d‘un cône et des mesures de paramètres de pénétration au cône dans les sols limoneux. Poulsen R., Nielsen B.N., Ibsen L.B.

603

Sampling method and pore water pressure measurement in the great depth (-400m) Méthode de mesure de pression interstitielle de l’eau d’échantillonnage en grande profondeur (– 400m) Rito F., Emura T.

607

Une méthode de classification de la sensibilité des sols au moyen du piézocône Soils sensibility classification method from piezocone data Serratrice J.-F.

611

Correction of soil design parameters for the calculation of the foundation based on the results of barrettes static load test Correction des paramètres de conception du sol pour le calcul sur la base des résultats de test de barrettes de charge statique Shulyatiev O., Dzagov A., Bokov I., Shuliatev S.

615

Characterization and Settlement Modeling of Deep Inert Debris Fills Caractérisation et modélisation du tassement de dépôts épais de gravats inertes Somasundaram S., Khilnani K., Shenthan T., Irvine J.

619

Site Characterization for the HZM Immersed Tunnel Caractérisation du site pour le tunnel immergé HZM Steenfelt J.S., Yding S., Rosborg A., Hansen J.G., Yu R.

623

Controversial and Contradictory Evaluations in Analyses of Ground Vibrations from Pile Driving Évaluations controversées et contradictoires dans l’analyse des vibrations de terre par suite de l’enfoncement de pieux Svinkin M.R.

629

CPT/PCPT- Based Organic Material Profiling Matière organique - Le profilage basé sur le CPT/PCPT Tümay M.T., Hatipkarasulu Y., Marx E.R., Cotton B.

633

Geotechnical Challenge for Total Cost Reduction related to Construction of Connecting Bridge with Pile Foundations Défi géotechnique pour la réduction totale des coûts liés à la construction du pont de liaison avec les fondations sur pieux Yasufuku N., Ochiai H., Maeda Y. Dynamic CBR as a method of embankment compaction assessment Dynamique CBR comme une méthode d’évaluation de compactage du remblai Zabielska-Adamska K., Sulewska M.J.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Technical ComitteeC103 Numerical Methods in Geomechanics Comité technique 103 Méthodes numériques en géomécanique General Report of TC103 Numerical Methods Rapport général du TC103 Méthodes numériques Chau K.T.

647

Equivalent pier theory for piled raft design Équivalence de la théorie de la jetée pour la conception de radeau empilé Balakumar V., Huang M., Oh E., Balasubramaniam A.S.

655

Interprétation d’essais d’extraction de renforcements métalliques haute adhérence dans un massif en Terre Armée® soumis à un chargement dynamique cyclique Interpretation of pullout tests of high adherence steel reinforcements in a Reinforced Earth® structure under a cyclic dynamic loadin Bennani Y., Soyez L., Freitag N.

659

On non-coaxial stress-dilatancy theories Sur les théories de non co-axialité contrainte/dilatance Biru A., Benz T.

663

On the geometry of plastic potential surfaces and isochoric stress paths Sur la géométrie des surfaces potentielles plastiques et des chemins de contraintes isochores Biru A., Benz T., Nordal S.

667

Modélisations de l’interaction sol-pieux pour le calcul d’impédances dynamiques Numerical modelling of soil-pile interaction and evaluation of dynamic impedances Breugnot A., Allagnat D., Baguelin F., Schlosser F., Osmani E., Servant C.

673

Validation of geotechnical finite element analysis Validation d’analyse par éléments finis pour la géotechnique Brinkgreve R.B.J., Engin E.

677

Evaluation of the efficiency of different ground improvement techniques Évaluation de l’efficacité des différentes techniques d’amélioration des sols Bryson S., El Naggar H.

683

Large deformation and post-failure simulations of segmental retaining walls using mesh-free method (SPH) Simulations de grandes déformations et post-rupture des murs de soutènement segmentaires utilisant la méthode des mailles-libres (SPH) Bui H.H., Kodikara J.A, Pathegama R., Bouazza A., Haque A.

687

Comparative Study on EQWEAP Analysis with 2D/3D FE Solutions Étude comparative sur l’analyse EQWEAP avec des solutions 2D/3D FE Chang D.-W., Wang Y.-C., Wu W.-L., Chin C.-T.

691

Large-Scale Geotechnical Finite Element Analysis on Desktop PCs Analyse par éléments finis de problèmes géotechniques de grandes dimensions sur ordinateur de bureau Chaudhary K.B., Phoon K.K., Toh K.C.

695

Calibration of a modified hardening soil model for kakiritic rocks Étalonnage d’un modèle modifié d’écrouissage des sols pour les roches kakiritiques Dong W., Anagnostou G.

699

Numerical investigations of shear strain localization in an elasto-plastic Cosserat material Investigations numériques sur les déformations en cisaillement dans un matériau élastoplastique de type Cosserat Ebrahimian B., Noorzad A.

703

Effect of Excavation-induced Movements on Adjacent Piles Effets des mouvements causés par une excavation sur les pieux voisins Elkady T.

707

Finite Element Modelling of D-wall Supported Excavations Modèle elément finis d’excavations soutenues par parois moulée Everaars M.J.C., Peters M.G.J.M.

711

3D simulation of overtopping erosion on embankments by shallow-water approximation Simulation en 3D d’une érosion par débordement sur des remblais, avec approximation en eau peu profonde Fujisawa K., Murakami A.

715

Numerical Investigations on Vibratory Sheet Piling in Embankments using a Multi-Phase Material Études numériques des effets de vibrofonçage sur les berges en utilisant une approche multi-phasique Hamann T., Grabe J.

719

Combined computational-experimental Laboratory Testing for Soil Behavior Modeling Combinaison d’essais numériques et expérimentaux pour la modélisation du comportement des sols Hashash Y.M.A., Asmar R., Moon S.

723

X

Contents / Table des matières

Numerical analysis on prediction for residual deformation of earth structure using rigid plastic dynamic deformation analysis Étude numérique pour prévoir la déformation résiduelle d’un ouvrage en terre à l’aide de l’analyse de la déformation dynamique rigide plastique Hoshina T., Isobe K.

727

Undrained bearing capacity of spatially random clays by finite elements and limit analysis Capacité portante des argiles non drainées des champs aléatoires par éléments finis et analyse limite Huang J., Lyamin A.V., Griffiths D.V., Sloan S.W., Krabbenhoft K., Fenton G.A.

731

On the use of waste rock inclusions to improve the performance of tailings impoundments Sur l’utilisation d’inclusions de roches stériles pour améliorer la performance des parcs à résidus miniers James M., Aubertin M., Bussière B.

735

Numerical modelling and control of seawater intrusion in coastal aquifers Modélisation numérique et contrôle des intrusions d’eau de mer dans les aquifères côtiers Javadi A.A., Hussain M.S., Abd-Elhamid H.F., Sherif M.M.

739

Computer Simulation of Levee’s Erosion and Overtopping Simulation numérique de l’érosion et de la surverse de digues Kamalzare M., Zimmie T.F., Han T.S., McMullan M., Cutler B., Franklin W.R.

743

Using 3D numerical solutions for the simplified modelling of interaction of soil and elongated structures Utilisation de solutions 3D numériques pour la modélisation simplifiée de l’interaction des sols et des structures allongées Kholmyansky M.L., Sheynin V.I.

747

3D Dynamic Numerical Modeling for Soil-Pile-Structure Interaction in Centrifuge Tests Modélisation numérique dynamique en 3D de l’interaction sol-pieu en centrifugeuse Kwon S.-Y., Kim M.-M., Kim S.-H., Choi J.-I

751

Two methods for estimating excess pore pressure in LEM Deux méthodes pour estimer l’excès de pression interstitielle Lehtonen V., Länsivaara T.

755

Comparison of 3D Finite Element Slopes Stability with 3D Limit Equilibrium Analysis Comparaison de la stabilité des éléments 3D pente finie avec l’analyse limite d’équilibre 3D Lu H.H., Xu L.M., Fredlund M.D., Fredlund D.G.

759

Modelling of soil-structure interaction for seismic analyses of the Izmit Bay Bridge Modélisation de l’interaction sol-structure pour l’analyse sismique du pont de la baie d’Izmit Lyngs J. H., Kasper T., Bertelsen K.S.

763

Numerical Analysis of a Tunnel Intersection Analyse numérique de l’intersection de tunnels Mayoral J.M., Román-de la Sancha A., Osorio L., Martínez S.

769

Numerical Evaluation of the Behavior of Reinforced Soil Retaining Walls Simulation numérique du comportement de murs de soutènement en sol renforcé Mirmoradi S.H., Ehrlich M.

773

Application of Genetic Algorithms with Hill Climbing Procedure to a Constitutive Model for Hard Soils and Soft Rocks Application des algorithmes génétiques avec la méthode de gradient à un modèle constitutif pour sols durs et roches tendres Pereira C., Caldeira L., Maranha das Neves E., Cardoso R.

777

Analytically and experimentally based resistance factors for “full-flow” penetrometers Résistance-facteurs pour “full flow” pénétromètres, basé sur résultats analytiques et expérimentaux Pinkert S., Klar A.

781

Analysis of ettringite attack to stabilized railway bases and embankments Analyse de l’attaque chimique par ettringite de remblais et plateformes ferroviaires stabilisées Ramon A., Alonso E.E.

785

The influence of buildings and ground stratification on tunnel lining loads using finite element method L’influence des bâtiments et de la stratification du sol sur les charges de revêtement du tunnel utilisant la méthode d’éléments finis Rezaei A.H., Katebi H., Hajialilue-Bonab M., Hosseini B.

789

Numerical Investigation of The Mobilization of Active Earth Pressure on Retaining Walls Enquête numérique de la mobilisation de la pression de la terre active sur les murs de retenue Sadrekarimi A., Damavandinejad Monfared S.

793

Artificial intelligence for modeling load-settlement response of axially loaded (steel) driven piles Application de l’intelligence artificielle à la modélisation de la courbe effort-tassement des pieux battus (en acier) soumis à un chargement axial Shahin M.A.

797

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

A visco-elasto-plastic multi-surface cyclic model Un modèle visco-élastoplastique de répétition de multi-surface Siddiquee, S.A., Islam K.

801

The design and construction of temporary works for Limerick Immersed Tube Tunnel Design et construction des travaux temporaires du tunnel-tube immergé de Limerick Smith A.K.C., Thorup O., Hudson J.

805

The application of the Iwan soil model on a deep excavation L’application du modèle de sol d’Iwan sur une excavation profonde Sokolić I., Szavits-Nossan A.

809

Numerical modelling of desiccation crack induced permeability Modélisation numérique de la perméabilité induite par la fissuration des sols Stirling R.A., Davie C.T., Glendinning S.

813

The tip resistance in layered soils during static penetration La résistance en pointe dans les sols stratifiés pendant une pénétration statique Sturm H.

817

Measured and Simulated Interactions between Kenaf Geogrid Limited Life Geosynthetics (LLGs) and Silty Sand Backfill Interactions mesurées et simulées entre kénaf géogrille limitée Géosynthétiques vie (LLGs) et de remblai de sable limoneux Tanchaisawat T., Bergado D.T., Artidteang S. Interaction between structures and compressible subsoils considered in light of soil mechanics and structural mechanics Étude de l’interaction sol- structures à la lumière de la mécanique des sols et de la mécanique des stuctures Ulitsky V.M., Shashkin A.G., Shashkin K.G., Vasenin V.A., Lisyuk M.B., Dashko R.E.

821

825

Rapid Drawdown Analysis using Strength Reduction Analyse d’abaissement rapide utilisant la force de réduction VandenBerge D.R., Duncan J.M., Brandon T.L.

829

Validation of computational liquefaction in plane strain Validation de liquéfaction simulée en déformation plane Wanatowski D., Shuttle D.A., Jefferies M.G.

833

Analysis of Ultimate Bearing Capacity of Single Pile Using the Artificial Neural Networks Approach A Case Study Analyse de la capacité portante ultime d’un pieu unique à l’aide de la méthode des réseaux de neurones artificiels : une étude de cas Wardani S.P.R., Surjandari N.S., Jajaputra A.A. Simulation of Delayed Failure in Naturally Deposited Clay Ground by Soil-water Coupled Finite Deformation Analysis Taking Inertial Forces into Consideration Simulation de rupture différée d’un sol d’argile naturelle sédimentaire à l’aide de l’analyse des déformations finies de squelette couplé eau-sol en tenant compte de la force d’inertie Yamada S., Noda T.

837

841

An elastic-viscous-plastic modeling of time-dependent behaviors of overconsolidated clays Un modèle élasto-visco-plastique pour les argiles surconsolidés Yao Y.P., Kong L.M.

845

Failure Modes for Geosynthetic Reinforced Column Supported (GRCS) Les modèles de rupture de massifs renforcés par colonnes sol-ciment et géosynthétiques (GRCS) Yapage N.N.S., Liyanapathirana D.S., Leo C.J.

849

The Material Point Method: A promising computational tool in Geotechnics La méthode du point matériel : un outil prometteur de calcul en géotechnique Yerro A., Alonso E., Pinyol N.

853

Development of excess pore-water pressure in thawing process of frozen subgrade soils: Based on analytical solutions and finite element method. Dégel des sols et variation de la pression d’eau interstitielle: application de méthodes analytiques et des éléments finis Yesuf G.Y., Hoff I., Vaslestad J. Prediction of stress and strain for the seabed and production well during methane hydrate exploitation in turbidite reservoir Prédiction de stress et déformation pour le fond de la mer et de puits pendant l’exploitation d’hydrate de méthane dans le réservoir du turbidité Yoneda J.

XII

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Contents / Table des matières

Technical Committee 104 Physical Modelling in Geotechnics Comité technique 104 Modélisation physique en géotechnique General Report for TC104 - Physical Modelling in Geotechnics Rapport général du TC104 - Modélisation physique en géotechnique White D.J., Gaudin C., Take W.A.

867

Centrifuge model tests on foundation on geosynthetic reinforced slope Essais en centrifugeuse d’une fondation sur une pente renforcée par géosynthétique Aklik P., Wu W.

875

Loading behaviour of flexible raft foundations in full scale and centrifuge models Comportement de radiers flexibles dans des essais grandeur nature et en centrifugeuse Arnold A., Laue J.

879

Investigation on the dynamic properties of saturated sand-tire chips mixture by shaking table Étude des propriétés dynamiques d’un mélange de sable saturé et de chute de pneumatiques sur table vibrante Bahadori H., Manafi S.

883

The influence of the g-level for anchor tests in sand L’influence du niveau de g pour les tests d’ancrage en sable Bezuijen A., Zwaan R., Lottum van H.

887

An experimental study on the consolidation of soft clayey soils using electrochemical methods Étude expérimentale de la consolidation des argiles molles avec des méthodes électrochimiques Cardoso R., Nogueira Santos J.

891

Variation of Friction Angle and Dilatancy For Anisotropic Cohesionless Soils Variations de l’angle de Frottement et de la Dilatance pour les Sols Anisotropes Sans Cohésion Cinicioglu O., Abadkon A., Altunbas A., Abzal M.

895

Centrifuge Modeling of Seismic Soil-Structure-Interaction and Lateral Earth Pressures for Large Near-Surface Underground Structures Modélisation en centrifugeuse de l’Interaction sol-structure sismique et des pressions de terre latérales pour les grands ouvrages souterrains proches de la surface Dashti S., Hushmand A., Ghayoomi M., McCartney J.S., Zhang M., Hushmand B., Mokarram N., Bastani A., Davis C., Lee Y., Hu J. Evaluation of Seismic Earth Pressure Reduction using EPS Geofoam Évaluation de la réduction de la poussée sismique en utilisant du Polystyrène Expansé Dave T.N., Dasaka S.M., Khan N., Murali Krishna A. Analysis of an adaptive foundation system for embankments on soft soils by means of physical and numerical modelling Analyse d’un système de fondation adaptatif pour les remblais sur sols compressibles par modélisation physique et numérique Detert O., Alexiew D., Schanz T., König D.

899

903

907

Reliability analysis of empirical predictive models for earthquake-induced sliding displacements of slopes Analyse de fiabilité des modèles empiriques de prédiction des déplacements sismiques de pentes Fotopoulou S., Pitilakis K.

911

Development of pore water pressure around a stone column Développement des pressions interstitielles autour d’une colonne ballastée Gautray J., Laue J., Springman S.M., Almeida M.

915

Large scale 1-g shake table model test on the response of a stiff pile group to liquefaction induced lateral spreading 919 Réponse d’un groupe de 3 × 3 pieux rigides sous l’action d’un écoulement latéral induit par liquéfaction étudié à grande échelle sur table vibrante Haeri S.M., Kavand A., Asefzadeh A., Rahmani I. Dynamic centrifugal model test for unsaturated embankments considering seepage flow and the numerical analysis Expérimentation en centrifugeuse et modélisation numérique de la réponse aux séismes de remblais non saturés en prenant en compte l’écoulement Higo Y., Oka F., Kimoto S., Kinugawa T., Lee C.-W., Doi T.

923

Développement d’un modèle réduit tridimensionnel du renforcement des sols par inclusions rigides Development of a three-dimensional small scale model to simulate soil improvement by rigid piles Houda M., Jenck O., Emeriault F., Briançon L., Gotteland Ph.

927

Full-scale field validation of innovative dike monitoring systems Validation de systèmes de surveillance innovants pour digues à grande échelle Koelewijn A.R., Vries (de) G., Lottum van H.

931

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Physical modeling of the vibration mitigation by an isolating screen Modélisation physique de l’atténuation des vibrations par un écran isolant Masoumi H., Vanhonacker P. The Drained Strength of Soft Clays with Partially Penetrating Sand Columns at Different Area Replacement Ratios La résistance drainée des argiles molles avec des colonnes de sable pénétrant partiellement à différents taux de remplacement Najjar S., Sadek S., Bou Lattouf H. Physical modeling of arch action in undercut slopes with actual engineering practice to Mae Moh open-pit mine of Thailand Modélisation physique de l’effet de voûte dans les pentes en déblai en suivant la pratique de l’ingénieur pour la mine à ciel ouvert à Mae Moh en Thaïlande Pipatpongsa T., Khosravi M.H., Takemura J.

935

939

943

Critical State Modelling of Soil-Structure Interface for Advanced Design Modélisation à l’état critique d’interface sol-structure pour la conception avancée Sarma D., Sarma M.D

947

A Study on the Influence of Ground Water Level on Foundation Settlement in Cohesionless Soil Étude de l’influence de la variation du niveau d’eau sur le tassement des fondations superficielles reposant sur sol granulaire Shahriar M.A., Sivakugan N., Urquhart A., Tapiolas M., Das B.M.

953

Water injection aided pile jacking centrifuge experiments in sand Essais en centrifugeuse d’installation de pieux vérinés dans le sable avec injection d’eau Shepley P., Bolton M.D.

957

Shear Behaviour of Rock Joints under CNS Boundary Conditions Comportement en cisaillement de joints rocheux en condition de raideur normale constante Shrivastava A.K., Rao K.S.

961

Experimental study on compaction grouting method for liquefiable soil using centrifuge test and X-ray tomography Etude expérimentale sur la CPG pour le sol liquéfiable par centrifugation et tomographie à rayons X Takano D., Morikawa Y., Nishimura S., Takehana K.

965

A model study of strains under footings supported by floating and end-bearing granular columns Une étude sur modèle réduit des contraintes sous semelles isolées reposant sur des colonnes granulaires flottantes et encastrées Tekin M., Ergun M.U.

969

Modélisation physique du blocage d’un écoulement d’eau dans un sol par injection d’un produit de colmatage Physical modelling of blocking phenomenon, by injection of a clogging product, of water flow through soils Truong Q.Q., Dupla J.-C., Canou J., Chevalier C., Chopin M., Fry J.J.

973

Hydraulic conductivity and small-strain stiffness of a cement-bentonite sample exposed to sulphates Conductivité hydraulique et module de cisaillement initial d’un échantillon de ciment-bentonite exposé aux sulfates Verástegui-Flores R.D., Di Emidio G., Bezuijen A.

977

Centrifuge modelling of bored piles in sands Modélisation en centrifugeuse de pieux forés dans le sable Williamson M.G., Elshafie M.Z.E.B., Mair R.J.

981

Stability and performance of ground improvement using geocell mattresses under extreme weather La stabilité et les performances de l’amélioration du sol en utilisant des matelas géocellules dans des conditions météorologiques extrêmes Xu Y., Wang J.P.

985

Technical committee 105 Geo-Mechanics from Micro to Macro Comité technique 105 Géomécanique micro-macro General Report of TC 105 - Geomechanics through the scales  Rapport général du TC 105 - La géomécanique à travers les échelles Viggiani G.

991

Un rêve devenu réalité : explorer une bande de cisaillement à l’échelle des grains Grain-scale experimental investigation of shear banding in sand Andò E., Desrues J., Bésuelle P., Viggiani G., Hall S.

999

Modelling crushing of granular materials as a poly-disperse mixture Modélisation de la fracturation des matériaux granulaires comme un mélange poli-disperse Caicedo B., Ocampo M., Vallejo L.

XIV

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Contents / Table des matières

Behaviour of a compacted silty sand under constant water content shearing Comportement d’un sable limoneux compacté sous cisaillement à teneur en eau constante Heitor A., Rujikiatkamjorn C., Indraratna B.

1007

Shear Strength and Deformation of Methane Hydrate Bearing Sand with Fines Résistance au cisaillement et déformation des sables avec des fines contenant de l’hydrate de méthane Hyodo M., Yoshimoto N., Kato A., Yoneda J.

1011

A Simplified Contact Model for Sandy Grains Cemented with Methane Hydrate Un modèle simplifié pour les contacts entre grains de sable cimentés par hydrates de méthane Jiang M., Liu F., Zhu F., Xiao Y.

1015

Macro- and micro-FE modelling of wellbore damage due to drilling and coring processes Modélisation par les éléments finis aux échelles micro et macro de l’endommagement dû au forage et au carrotage Khoa H.D.V., Grande L., Jostad H.P.

1019

Three dimensional discrete element simulation of trapdoor unloading and gravity flow of sandy granular material Simulation tridimensionnelle par les éléments distincts du débit de décharge et d’écoulement gravitaire du matériau granulaire sableux Kikkawa N., Itoh K., Toyosawa Y., Pender M.J., Orense R.P.

1023

Microstructural changes leading to chemically enhanced drainage Modifications de microstructure entraînant un drainage chimiquement amélioré Minder P., Puzrin A.M.

1027

Discrete Element Method Study of Shear Wave Propagation in Granular Soil Étude de la propagation des ondes de cisaillement dans un sol granuleux par la méthode des éléments discrets Ning Z., Evans T.M.

1031

Microscopic observation on compacted sandy soil using micro-focus X-ray CT Observation microscopique par micro-tomographie à rayons X de sables compactés Otani J., Mukunoki T., Takano D., Chevalier B.

1035

Study of relative permeability variation during unsteady flow in saturated reservoir rock using Lattice Boltzmann method Étude de la variation de la perméabilité relative au cours d’écoulement transitoire dans une roche réservoir saturée en utilisant la méthode des réseaux de Boltzmann Pak A., Sheikh B.

1039

Uniform effective stress equation for soil mechanics Équation aux contraintes effectives uniformes pour la Mécanique des Sols Shao L.-T., Liu G., Guo X.-X.

1043

Particulate Modeling of Sand Slurry Flow Retardation Modélisation par les milieux granulaires de l’effet de retard de l’écoulement des boues résiduelles Tomac I., Gutierrez M.

1047

A Coupled Analysis of Fluid-Particle Interactions in Granular Soils Analyse couplée des interactions fluide-particules dans les sols granulaires Zhao J., Shan T.

1051

Experimental study of resilient modulus of unsaturated soil at different temperatures Etude expérimentale du module de résilience d’un sol non saturé à différentes températures Zhou C., Ng C.W.W.

1055

Technical committee 106 Unsaturated Soils Comité technique 106 Sols non saturés General Report of TC 106 - Unsaturated soils Rapport général du TC 106 - Sols non saturés Jommi C.

1061

A simple approach for predicting vertical movements of expansive soils using the mechanics of unsaturated soils Une approche simple pour prédire les mouvements verticaux des sols gonflants par la mécanique des sols non saturés Adem H.H., Vanapalli S.K. Étude de l’impact de l’hygrométrie sur la fissuration d’un sol gonflant Impact of the hygrometry on the swelling soil cracking Auvray R., Rosin-Paumier S., Abdallah A., Masrouri F.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

One-dimensional cracking model in clayey soils Modélisation unidimensionnel de la fissuration des sols argileux Ávila G., Ledesma A., Lloret A. Experimental Study on Effect of Initial Moisture Content on Compressive Property of Compacted Loess Like Silt Étude expérimentale des caractéristiques de compression des lœss compactés Bai X., Yang J., Ma F.

1077

1081

Evolution of microstructure during desiccation of oil sands mature fine tailings Évolution de la microstructure en séchage des résidus de sables bitumineux Bajwa T, Simms P.

1085

Evaluation of void ratio and elastic modulus of unsaturated soil using elastic waves Évaluation de l’indice des vides et du module élastique d’un sol non saturé en utilisant les ondes élastiques Byun Y.H., Lee J.S., Cho S.H., Yoon H.K.

1089

Evaluation Curves SWCC for Tropical Peruvian Soils Évaluation des courbes de rétention d’eau SWCC pour les sols tropicaux péruviens Carrillo-Gil A., Carrillo-Acevedo A.

1093

Étude par la méthode des éléments finis du comportement des remblais en sols fins compactés Finite element analysis of embankments in fine compacted soils Droniuc N.

1097

Comportement des sols gonflants lors de l’humidification et du séchage Behavior of swelling soil under cyclic wetting and drying Ejjaaouani H., Shakhirev V., Magnan J.-P., Bensallam S.

1101

Numerical study of damage in unsaturated bentonite with θ-stock finite element code Étude numérique d’endommagement pour les milieux poreux non saturés avec le code des éléments finis θ-stock Fathalikhani M., Gatmiri B.

1105

Combination of Shrinkage Curve and Soil-Water Characteristic Curves for Soils that Undergo Volume Change as Soil Suction is Increased Combinaison des courbes de retrait et des courbes des propriétés hydriques des sols pour les sols subissant un changement de volume avec une augmentation de la succion Fredlund D.G., Zhang F.

1109

Small-strain shear modulus and shear strength of an unsaturated clayey sand Module de cisaillement en petites déformations et la résistance au cisaillement d’un sable argileux non saturé Georgetti G.B., Vilar O.M., Rodrigues R.A.

1113

Étude de la stabilité des pentes non saturées sous les effets de l’infiltration prenant en compte la végétation Study of the stability of unsaturated slopes under the effects of infiltration taking into account the vegetation Hemmati S., Modaressi A.

1117

Rainfall-induced collapse of old railway embankments in Norway Influence des precipitations sur l’instabilité d’anciens remblais ferroviaires en Norvège Heyerdahl H., Høydal Ø., Nadim F., Kalsnes B.G., Børsting T.

1121

Dynamic shear modulus and damping of compacted silty sand via suction-controlled resonant column testing Propriétés dynamiques d’un sable limoneux par des tests en colonne de résonance sous aspiration contrôlée. Hoyos L.R., Cruz J.A., Puppala A.J., Douglas W.A., Suescún E.A.

1125

Expression of mechanical characteristics in compacted soil with soil/water/air coupled F.E. simulation Expression des caractéristiques mécaniques des sols compactés par une simulation couplée sol/eau/air par éléments finis Kawai K., Iizuka A., Kanazawa S.

1129

A Geotechnical Countermeasure for Combating Desertification Une mesure géotechnique pour lutter contre la désertification Liu Q., Yasufuku N.

1133

Extension of measurement range of dew-point potentiometer and evaporation method Extension de gamme de mesure de potentiomètre de point de rosée et méthode d’évaporation Maček M., Smolar J., Petkovšek A.

1137

Field capacity and moisture loss during active deposition on Tailings Dams Capacité au champ et perte d’humidité pendant le dépôt actif des résidus MacRobert C.

1143

Effet du retrait du sol sur une maison expérimentale Effects of soil shrinkage on an experimental house Makki L., Bourgeois E., Burlon S., Magnan J.-P., Duc M.

1147

Hydro-mechanical properties of lime-treated London Clay Propriétés hydromécaniques de l’argile de Londres traitée à la chaux Mavroulidou M., Zhang X., Kichou Z., Gunn M.J.

1151

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Contents / Table des matières

Influence of initial water content on the water retention behaviour of a sandy clay soil Influence de la teneur en eau initiale sur le comportement de rétention d’eau d’une argile sableuse Mendes J., Toll D.G.

1155

Climate Change Effects on Expansive Soil Movements Les effets du changement climatique sur les mouvements d’un sol gonflant Mitchell P.W.

1159

Study on mechanism of two-phase flow in porous media using X-ray CT Image Analysis Etude sur le mécanisme de transfert biphasé dans les milieux poreux par l’imagerie aux rayons X Mukunoki T., Mikami K.

1163

Soil suction induced by grass and tree in an atmospheric-controlled plant room Succion du sol induite par l’herbe et l’arbre dans une chambre atmosphérique contrôlée Ng C.W.W., Leung A.K., Garg A., Woon K.X., Chu L.M., Hau B.C.H.

1167

Application of micro-porous membrane technology for measurement of soil-water characteristic curve Application de la technologie de membrane microporeuse pour la détermination de la courbe de rétention d’eau des sols Nishimura T.

1171

Determination of soil-water retention curve for a young residual soil using a small centrifuge Détermination de la courbe de rétention d’eau pour un sol résiduel jeune à l’aide d´une petite centrifugeuse Reis R.M., Saboya F., Tibana S., Marciano C.R., Ribeiro A.B., Sterck W.N., Avanzi E.D.

1175

Interpretation of the Effect of Compaction on the Mechanical Behavior of Embankment Materials Based on the Soil Skeleton Structure Concept Interprétation de l’effet de compactage sur le comportement mécanique des matériaux de remblai basée sur le concept de structure des sols Sakai T., Nakano M.

1179

Mechanisms of Strength Loss during Wetting and Drying of Pierre Shale Mécanismes de la perte de force pendant humidification et séchage de Pierre Shale Schaefer V.R., Birchmier M.A.

1183

Effect of confining stress on the transient hydration of unsaturated GCLs Effet de la contrainte de confinement sur l’hydratation transitoire de GCLs insaturés Siemens G.A., Take W.A., Rowe R.K., Brachman R.

1187

Soil chart, new evaluation method of the swelling-shrinkage potential, applied to the Bahlui’s clay stabilized with cement L’ empreinte du sol, une nouvelle méthode d’évaluation du potentiel de gonflement, appliquée à l’argile de Bahlui stabilisée avec du ciment. Stanciu A., Aniculaesi M., Lungu I. Measurement of Unsaturated Ground Hydraulic Properties using a Dynamic State Soil Moisture Distribution Model Mise en œuvre de l’évaluation d’une mesure des propriétés hydrauliques d’un sol non saturé par un modèle dynamique de distribution de l’humidité Sugii T., Yamada K., Asano N., Yamada Y.

1191

1195

New devices for water content measurement Les appareils nouveaux pour la mesure de la teneur en eau Toll D.G., Hassan A.A., King J.M., Asquith J.D.

1199

A simplified model for collapse using suction controlled tests Un modèle simplifié d’effondrement, basée sur des essais de succion controlée Vázquez M., Justo de J.L., Durand P.

1203

Critical State for Unsaturated Soils and Steady State of Thermodynamic Process Etat critique de s sols non saturés et état stable thermodynamique Zhao C.G., Li J., Cai G.Q., Liu Y.

1207

Technical committee 202 Transportation Geotechnics Comité technique 202 Géotechnique des transports General Report TC202 - Transportation Geotechnics Rapport général du TC202 - Géotechnique pour les infrastructures de transport Indraratna B., Correia A. Five years of Impact Compaction in Europe – successful implementation of an innovative compaction technique based on fundamental research and field experiments Cinq ans de compactage par impact en Europe – mise en œuvre avec succès d’une technique de compactage novatrice basée sur la recherche fondamentale et expériences sur le terrain Adam D., Paulmichl I., Adam C., Falkner F.-J.

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Développement d’un modèle non linéaire de la voie ferrée ballastée Development of a non-linear ballasted railway track model Alves Fernandes V., Costa d’Aguiar S., Lopez-Caballero F.

1229

LGV EST lot 41 : tassements calculés puis mesurés sous remblais LGV EST section 41 : measured and calculated settlements under embankments Boutonnier L., Hajouai F., Bacar Fadhuli N., Gandille D.

1233

Recent developments in pavement foundation design Développements récents dans la conception des fondations des chaussées Brown S.F., Thom N.H.

1237

Deformation Performance and Stability Control of Multi-stage Embankments in Ireland Performance en déformation et contrôle de stabilité de remblais construits par étapes en Irlande Buggy F.J.

1241

Renforcement de plates-formes ferroviaires par colonnes de soil mixing réalisées sans enlever la voie Railways platforms reinforced by soil-mixing columns without track removing Calon N., Robinet A., Costa d’Aguiar S., Briançon L., Cojean C., Mosser J.-F.

1245

Analysis of the influence of soft soil depth on the subgrade capacity for flexible pavements Analyse de l’influence de la profondeur d’un sol mou sur la capacité portante pour les chaussées souples. Carvajal E., Romana M.

1249

The Use of Jet Grouting to Enhance Stability of Bermed Excavation L’utilisation de Jet Grouting pour améliorer la stabilité d’une excavation avec risbermes Cheuk J.C.Y., Lai A.W.L., Cheung C.K.W., Man V.K.W., So A.K.O.

1255

The geotechnical analysis corresponding to the high road embankments close to a bridge L’analyse géotechnique correspondant aux remblais routiers de grande hauteur à proximité d’un pont Chirica A., Vintila D., Tenea D.

1259

Applicability of the Geogauge, P-FWD and DCP for compaction control Étude des conditions d’application du Geogauge, DP et PDL dans le contrôle du compactage Conde M.C., Lopes M.G., Caldeira L., Bilé Serra J.

1263

Equilibrium models for arching in basal reinforced piled embankments Modèles d’équilibre par effet voute pour l’amélioration des sols de fondation par inclusions rigides Eekelen van S.J.M., Bezuijen A.

1267

Prise en compte des effets de la surconsolidation dans la stabilité des talus Consideration of Overconsolidation in slopes stability Guerpillon Y., Virollet M.

1271

Effects of ballast thickness and tie-tamper repair on settlement characteristics of railway ballasted tracks Les effets de l’épaisseur de ballast et de la réparation de lien-bourreur sur le tassement des voies chemin de fer Hayano K., Ishii K., Muramoto K.

1275

Mécanismes de transfert de charges dans les remblais sur cavités renforcés par géotextiles : approches expérimentales et numériques Load transfer mechanisms in geotextile-reinforced embankments overlying voids: experimental and numerical approaches Huckert A., Garcin P., Villard P., Briançon L., Auray G. Performance Assessment of Synthetic Shock Mats and Grids in the Improvement of Ballasted Tracks Évaluation de la performance des nappes synthétiques à effet d’amortissement et des géogrilles dans l’amélioration des plates-formes ferroviaires ballastées Indraratna B., Nimbalkar S., Rujikiatkamjorn C., Neville T., Christie D. Effect Evaluation of Freeze-Thaw on Deformation-Strength Properties of Granular Base Course Material in Pavement Évaluation des effets de gel-dégel sur les propriétés de résistance à la déformation des matériaux granulaires de couche de base des chaussées Ishikawa T., Zhang Y., Kawabata S., Kameyama S., Tokoro T., Ono T.

1279

1283

1287

Long-term performance of preloaded road embankment Comportement à long terme d’un remblai routier préchargé Islam M.N., Gnanendran C.T., Sivakumar S.T., Karim M.R.

1291

Probabilistic Settlement Analysis For The Botlek Lifting Bridge Design Analyse probabiliste de tassement pour la conception du pont levant Botlek Jacobse J.A., Nehal R.S., Rijneveld B., Bouwmeester D.

1295

Ground improvement methods for the construction of the federal road B 176 on a new elevated dump in the brown coal region of MIBRAG Méthodes d’amélioration de sols pour la construction de la route nationale B 176 traversant un remblai récent d’une mine de lignite de MIBRAG Kirstein J.F., Ahner C., Uhlemann S., Uhlich P., Röder K.

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Contents / Table des matières

Model tests on settlement behaviour of ballasts subjected to sand intrusion and tie tamping application Tests de modélisation sur le comportement en tassement des ballasts sujets à l’intrusion de sable et au bourrage Kumara J., Hayano K.

1305

Assessing the Effectiveness of Rolling Dynamic Compaction Évaluation de l’efficacité du compactage dynamique roulant Kuo Y.L., Jaksa M.B., Scott B.T., Bradley A.C., Power C.N., Crisp A.C., Jiang J.H.

1309

Determination of distribution of modulus of subgrade reaction Détermination de la distribution du module de réaction d’un sol de fondation Larkela A., Mengelt M., Stapelfeldt T.

1313

Stability improvement methods for soft clays in a railway environment Méthodes d’amélioration de la stabilité des argiles moles sous remblai de chemin de fer Mansikkamäki J., Länsivaara T.

1317

Effect of wetting- drying cycles on CBR values of silty subgrade soil of Karaj railway Effet des cycles d’humidification et séchage sur les valeurs CBR des sols de limoneux de fondation de la voie ferrée Karaj Moayed R.Z., Lahiji B.P., Daghigh Y.

1321

On the Permanent Deformation Behavior of Rail Road Pond Ash Subgrade Sur le comportement en déformation permanente d’une assise ferroviaire en cendres volantes de bassin Mohanty B., Chandra S.

1325

Evaluation of the Performance of Road Embankments over North Jakarta-Soft Soils Évaluation de la performance de remblais routiers sur les sols mous du Nord de Djakarta. Murjanto D., Rahadian H., Hendarto, Taufik R.

1329

Retrofit Technique for Asphalt Concrete Pavements after seismic damage Technique de réhabilitation pour chaussée en béton d’asphalte après dommage sismique Ohta H., Ishigaki T., Tatta N.

1333

Simultaneous interpretation of CPT/DMT tests to ground characterisation L’interprétation simultanée des essais CPT/DMT pour la caractérisation du sol Rabarijoely S., Garbulewski K.

1337

Modélisation numérique 3D d’un système de fondation d’un complexe immobilier 3D numerical modeling of a foundation system of a building complex Reynaud S., Allagnat D., Mazaré B., Julien T.

1341

Comportement du viaduc élevé de la ligne 12 du métro de la Ville de Mexico, autour de la Sierra de Santa Catarina Elevated Viaduct behavior of Metro Line 12 Mexico City in the nearness of the Santa Catarina Rodríguez G.L.B., Soria C.B.

1345

Influence of installation damage on the tensile strength of asphalt reinforcement products Influence de l’endommagement de mise en place sur la traction des produits de renforcement en asphalte Sakou Touole L., Thesseling B.

1349

Influence of Anti-freezing layer on the Frost Penetration Depth for Paved Road Design Influence d’une couche anti-gel sur la profondeur de pénétration du gel dans la conception des chaussées Shin E.C., Cho G.T., Lee J.S.

1353

Evaluation of roadbed potential damage induced by swelling/shrinkage of the subgrade Effet du retrait-gonflement des sols sur les structures de chaussées Simic D.

1357

The performance of shale as fill and embankment material for a trunk road in Ghana La performance du schiste comme matériau de remblai pour une route destinée au trafic de camions au Ghana Solomon K.M., Oddei J.K., Gawu S.K.

1361

Influence of Mechanical Indices for Soil Basement on Strength of Road Structure Influence des paramètres mécaniques de la couche de fondation sur la résistance d’une structure de chaussée Teltayev B.

1365

Design and performance of a jet grout retaining wall in a railway embankment on soft soil Dimensionnement et performance d’une paroi de soutènement réalisée à l’aide de la technique de jet grouting dans un remblai ferroviaire sur sol mou Verstraelen J., Maekelberg W., Lejeune C., De Clercq E., De Vos L.

1369

Laboratory characterization and model calibration of a cemented aggregate for application in transportation infrastructures Caractérisation en laboratoire et calibration d’un modèle d’agrégat cimenté pour une utilisation dans les infrastructures de transport Viana da Fonseca A., Rios S., Domingues A.M., Silva A., Fortunato E.

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Special Aspects for Building a Motorway on a 185 m Deep Dump Aspects particuliers pour construire une autoroute sur un remblai de comblement de 185 m Vogt N., Heyer D., Birle E., Vogt S., Dahmen D., Karcher C., Vinzelberg G., Eidam F.

1377

Performance verification of a geogrid mechanically stabilised layer Vérification de la performance d’une couche stabilisée mécaniquement par une géogrille Wayne M., Fraser I., Reall B., Kwon J.

1381

Characterization of Soil-Geosynthetic Interaction under Small Displacements Conditions Caractérisation de l’Interaction sol-géosynthétique sous des conditions de petits déplacements Zornberg J.G., Roodi G.H., Ferreira J., Gupta R.

1385

Technical committee 203 Earthquake Geotechnical Engineering and Associated Problems Comité technique 203 Géotechnique parasismique et problèmes associés 1st General Report for TC 203: Seismic response of soils, foundations and geotechnical structures 1er rapport général du TC 203 : Réponse sismique des sols, des fondations et des ouvrages géotechniques Semblat J.-F.

1391

2nd General Report for TC 203: Experimental characterization and analysis of soil behaviour under earthquake loads 1399 2e rapport général du TC 203 : Caractérisation expérimentale et analyse du comportement des sols sous chargement sismique Silvestri F. Analytical study of seismic slope behavior in a large-scale shaking table model test using FEM and MPM Étude analytique du comportement des pentes sismiques dans les essais de modèles de grandes dimensions sur table à secousses conformément aux méthodes FEM et MPM Abe K., Izawa J., Nakamura H., Kawai T., Nakamura S.

1407

Degradation of clay due to cyclic loadings and deformations La dégradation de l’argile due à des chargements et des déformations cycliques Åhnberg H., Larsson R., Holmén M.

1411

The effect of fines type on correlation between shear wave velocity and liquefaction resistance of sand containing fines L’effet du type amendes sur la corrélation entre la vitesse des ondes de cisaillement et de résistance à la liquéfaction du sable contenant des amendes Akbari-Paydar N., Ahmadi M.M.

1415

Dependency of nonuniform ground surface liquefaction damage on organization and slope of deep strata Non-uniformité des dommages de liquéfaction de la couche de surface due à la configuration des strates profondes et de l’inclinaison des strates Asaoka A., Nakai K.

1419

Seismic slope stability of earthen levees La stabilité sismique de pente de digues en terre Athanasopoulos-Zekkos A., Seed R.B.

1423

3D Numerical Analysis of a Suspension Bridge Anchor Block to Oblique-Slip Fault Movement Analyse numérique 3D d’un bloc d’ancrage de pont suspendu soumis à un mouvement oblique de glissement dû à une faille de rupture Avar B.B., Augustesen A.H., Kasper T., Steenfelt J.S.

1427

Seismic site effects in the city of Mendoza and surroundings (Argentina) Effets de site sismique dans la ville de Mendoza et les environs (Argentine) Barchiesi A.M., Mancipe-Herrera C.

1431

Liquefaction impact revisited L’impact de la liquéfaction revisité Barends F.B.J., Meijers P., Schenkeveld F.M., Weijers J.B.A.

1435

An experimental approach to evaluate shear modulus and damping ratio of granular material Une approche expérimentale pour évaluer le module de cisaillement et le taux d’amortissement du matériau granulaire Bolouri Bazaz J., Bolouri Bazaz H.R.

1439

Behavior of a multi-story building under seismic loads when taking into account the viscoplasticity of the soil base L’interaction entre les constructions du bâtiment sous charges sismiques tout en tenant compte de la viscoplasticité de la base du sol. Boyko I.P., Sakharov O.S., Sakharov V.O. Vers les métamatériaux sismiques Towards seismic metamaterials Brûlé S., Javelaud E., Guenneau S., Enoch S.

1443

1447

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Possibilities and limitations of the Prevost model for the modelling of cohesionless soil cyclic behaviour Possibilités et limitations du modèle de Prévost pour la modélisation du comportement cyclique des sols sans cohésion. Cerfontaine B., Charlier R., Collin F.

1451

On Seismic Performance and Load Capacities for Pile Design A propos des performances sismiques et les capacités de charge pour la conception de pieux Chang D.-W., Sung S.-H., Lee S.-M., Zhussupbekov A., Erlan Saparbek E.

1455

Challenges to the laboratory evaluation of field liquefaction resistance Les défis de l’évaluation en laboratoire de la résistance à la liquéfaction de terrain Coelho P.A.L.F., Azeiteiro R.J.N., Marques V.D., Santos L.M.A., Taborda D.M.G.

1459

Liquefaction Susceptibility of Loose Calcareous Sand of Northern Coast in Egypt La susceptibilité à la liquéfaction du sable calcaire lâche de la côte nord en Égypte Elmamlouk H., Salem M., Agaiby S.S.

1463

Seismic bearing capacity of strip footings near cohesive slopes using lower bound limit analysis Capacité portante séismique des fondations superficielles en bord des talus purement cohérents ; une évaluation par défaut suivant la méthode du calcul à la rupture Farzaneh O., Mofidi J., Askari F.

1467

Risk minimisation in construction of upstream tailings storage facilities based on in-situ testing Minimisation du risque sur base d’essais in situ lors de la construction de digues de stockage des résidus miniers par la méthode amont. Fourie A B., Palma J H., Villavicencio G., Espinace R.

1471

Dynamic soil-pile behavior in liquefiable sand overlaid with soft clay Dynamique sol-pieu comportement dans le sable liquéfiable recouvert d’argile molle Ghotbi S.M.A., Olyaei M., Yasrebi S.S., Mosallanejad M.

1475

Correlations between the shear wave velocity profile and the response spectrum based on SASW tests Corrélation entre le profil de vitesse d’ondes de coupe et le spectre de réponse basé sur l’essai SASW Gonzalez L., Pinilla C., Peredo V., Boroschek R.

1479

Methodological approach for the stability analysis of the Po river banks Méthodologie pour l’analyse de la stabilité des digues de la rivière Pô Gottardi G., Madiai C., Marchi M., Tonni L., Vannucchi G.

1483

Effect of Soil Plugging on Axial Capacity of Open-Ended Pipe Piles in Sands (manque traduction en français) Gudavalli S.R., Safaqah O., Seo H.

1487

Strain Response Envelopes for low cycle loading processes Enveloppe de réponse d´allongement pour chargements cycliques de basse intensité Hettler A., Danne St.

1491

Development of Map of Maximum Considered Earthquake Geometric Mean (MCEG) PGA for Earthquake Resistance Building Design in Indonesia Élaboration de la carte de moyenne géométrique du tremblement de terre maximum considéré (MCEG) PGA pour la conception antisismique des bâtiments en Indonésie Irsyam M., Asrurifak M., Ridwan M., Aldiamar F., Wayan Sengara I., Widiyantoro S., Triyoso W., Hilman D., Kertapati E., Meilano I., Suhardjono, Hendriyawan, Simatupang P.T., Muhammad I., Murjanto D., Hasan M.

1495

Study on long-term subsidence of soft clay due to 2007 Niigata Prefecture Chuetsu-Oki Earthquake Étude sur l’affaissement à long terme d’argile molle dû au tremblement de terre de la préfecture de Niigata Chuetsu-Oki en 2007 Isobe K., Ohtsuka S.

1499

Effect of stress anisotropy on cyclic behavior of dense sand with dynamic hollow cylinder apparatus Effet de l’anisotropie de contrainte sur le comportement cyclique du sable dense avec dynamique appareil cylindre creux Jafarzadeh F., Zamanian M.

1503

Impact of blast vibrations on the release of quick clay slides Impact des vibrations dues aux explosions sur les glissements de terrain dans les argiles sensibles Johansson J., Løvholt F., Andersen K.H., Madshus C., Aabøe R.

1507

Dynamic calculation for the dry closure of Almagrera tailings dam Calcul dynamique pour la fermeture à sec du barrage des stériles d’Almagrera Justo de J.L., Morales-Esteban A., Durand P., Vázquez-Boza M., Jiménez F.A., Rossi E.

1511

Recent developments in procedures for estimation of liquefaction potential of soils Développements récents des méthodes d’estimation du potentiel de liquéfaction des sols Katzenbach R., Clauss F., Rochée S.

1515

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Applying Earthquake Risk Analysis Methods to a Town in Hungary L’application des méthodes d’analyse du risque sismique dans le cas d’une ville de Hongrie Kegyes-Brassai O., Ray R.P.

1519

Ensuring Stability of Boards of Deep Ditches in Seismic Regions Assurer la stabilité des fossés profonds dans les régions sismiques Khomyakov V.A.

1523

Colonne à Module Mixte CMM® sous des sollicitations statiques et dynamiques : étude expérimentale Mixed Module Columns® under static and dynamic load – Experimental study. Lambert S., Santruckova H., Foray P., Flavigny E., Gotteland Ph.

1527

Évaluation de la réduction du risque de liquéfaction par des colonnes ballastées Valuation of liquefaction mitigation by stone columns Lambert S.

1531

Normalized Shear Modulus of Compacted Gravel Module de cisaillement normalisé des graviers compactés Liao T., Massoudi N., McHood M., Stokoe K.H., Jung M.J., Menq F.-Y.

1535

Dynamic Properties and Liquefaction Potential of a Sandy Soil Containing Silt Propriétés dynamiques et potentiel de liquéfaction d’un sol sablonneux contenant de la vase Mominul H.M., Alam M.J., Ansary M.A., Karim M.E.

1539

Seismic stability assessment of a steel plate fabricated column constructed on liquefiable grounds with different soil-layer profiles Évaluation de la stabilité sismique d’une colonne en plaques d’acier construite sur des sols liquéfiables avec différents profils sol-couche Nakai K., Xu B., Takaine T.

1543

A method of suppressing liquefaction using a solidification material and tension stiffeners Étude de base sur les méthodes de résistance à la liquéfaction, en utilisant des matériaux précontraints Nakamichi M., Sato K.

1547

Effects of Fines Content on Cyclic Shear Characteristics of Sand-Clay Mixtures Les effets de la teneur en fines sur les caractéristiques du cisaillement répété des mélanges de sable et argile Noda S., Hyodo M.

1551

Case study of the post-earthquake behavior of a CFRD dam Étude de cas sur le comportement post-sismique d’un barrage CFRD Núñez E.A. Sfriso O.

1555

Liquefaction characteristics of crushable pumice sand Caractéristiques de liquéfaction des sables de pierre ponce sensibles à l’écrasement Orense R.P., Pender M.J.

1559

Investigation of Reinforced Earth Structures Following the 2011 Tohoku Earthquake Etude des structures en Terre Armée suite au séisme de Tohoku de 2011 Otani Y., Takao K., Sakai S., Kimura T., Kuwano J., Freitag N., Sankey J.

1563

Accumulated Stress Based Model for Prediction of Residual Pore Pressure Étude et développement du modèle pour le pronostic sur l’excès de pression hydrostatique interstitielle causé par les contraintes accumulées Park D., Ahn J.-K.

1567

Pioneer application of a dynamic penetrometer and boroscope in archeological prospecting Application pionnière d’un pénétromètre dynamique et d’un boroscope dans la prospection archéologique Rangel-Núñez J.L., Barba L., Ovando E., Auvinet G., Ibarra-Razo E.

1571

Measuring and modeling the dynamic behavior of Danube Sands Mesure et modélisation du comportement dynamique des sables du Danube Ray R.P., Szilvágyi Z.

1575

Three-dimensional seismic active earth pressure coefficients using upper bound numerical limit analysis: a few preliminary results Coefficients de poussée tridimensionels séismiques déterminés avec une application numérique du theorème cinématique de l’analyse limite: quelques résultats préliminaires Santana T., Guerra N.M.C., Antão A.N., Vicente da Silva M. Modélisation 1D-3Composantes de la réponse sismique d’une colonne de sol multicouche à comportement non linéaire 1Directional-3Component seismic response modelling of a multilayer nonlinear soil profile. Santisi d’Avila M.P., Lenti L., Semblat J.-F. The behaviour of natural cohesive soils under dynamic excitations Le comportement des sols cohérentes naturelles sous excitations dynamiques Sas W., Szymański A., Gabryś K.

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Contents / Table des matières

Dynamic behavior of multi-arch culverts embankment considering the installation interval of consecutive arch culverts Comportement dynamique des terre-pleins à dalot multi-arche en fonction de l’intervalle entre les arches successives Sawamura Y., Kishida K., Kimura M. Méthode simplifiée de calcul d’une paroi sous séisme Simplified seismic wall stability analysis Serratrice J.-F.

1591

1595

Effect of Seismic Waves with Different Dominant Frequencies on the Delayed Failure Behavior of a Soil StructureGround System Effets des ondes sismiques de fréquence dominante différente sur le comportement de rupture retardée de structures en terre et de systèmes de sol Shimizu R., Yamada S. Shaking table test of large-scaled slope model subjected to horizontal and vertical seismic loading using E-Defense Tessai sur table à secousses de modèles de gros talus par une accèlèration vericale et horizontale par E-Dèfense Shinoda M., Nakajima S., Nakamura H., Kawai T., Nakamura S.

1599

1603

Stability analysis of earth dams under static and earthquake loadings using geosynthetics as a seepage barrier Analyse de stabilité des barrages en terre sous des charges statiques et sous séisme à l’aide de géosynthétiques comme une barrière d’infiltration Srivastava A., Sivakumar Babu G.L.

1607

Cyclic Loading Behavior of Saturated Sand with Different Fabrics Comportement du sable saturé avec des structures différentes sous chargement cyclique Sze H.Y., Yang J.

1611

Evaluation of effective parameters on soil layers seismic amplification ratios (A case study of Bam earthquake) Évaluation des paramètres effectifs sur les ratios d’amplification sismique des couches de sol (Une étude de cas de tremblement de terre de Bam) Tabatabaie S.H., Hassanlourad M., Yazdanparast M., Mohammadi A.

1615

Experimental study on lattice-shaped cement treatment method for liquefaction countermeasure Étude expérimentale d’un procédé d’anti-liquéfaction des sols au moyen d’un bâti en forme de treillage en béton Takahashi H., Morikawa Y., Iba H., Fukada H., Maruyama K., Takehana K.

1619

Shaking model tests on mitigation of liquefaction-induced ground flow by new configuration of embedded columns 1623 Essais sur table vibrante pour une attenuation de l’écoulement des sols du a la liquefaction par une nouvelle configuration de colonnes enterrees Takahashi N., Derakhshani A., Rasouli R., Towhata I., Yamada S. Structure-Soil Massif System Behavior Features Under Static & Dynamic Loads Les particularités du comportement du système edifice-sol avec des efforts statiques et dynamiques Taranov V.G., Aleksandrovych V.A., Luchkovskyi I. Ia., Plashchev S.A., Kornienko N.V., Areshkovych O.O.

1627

Pseudo static analysis considering strength softening in saturated clays during earthquakes L’analyse pseudo statique considérant la force de ramollissement dans l’argile saturée lors des tremblements de terre Tsai C.-C., Mejia L.H., Meymand P.

1631

Effectiveness of In-soil Seismic Isolation taking into account of Soil-Structure Interaction Efficacité d’ Isolement sismique dans le Sol tenant compte de l’interaction du Sol avec la Structure Tsatsis A.K., Anastasopoulos I.C., Gelagoti F.L., Kourkoulis R.S.

1635

The device of the bases and foundation in the conditions of weak soil and high seismic activity of the Republic of Tajikistan L’appareil des bases et de la fondation dans les conditions de faible sol et la haute activité sismique de la République du Tadjikistan Usmanov R.

1639

Foundation conditions analysis for some eolian power units corresponding to the seismic loads influence Analyse des fondations pour certaines unités d’éoliennes sous chargement sismique Vintila D., Tenea D., Chirica A.

1643

Performance-based Evaluation of Saturated Loess Ground Liquefaction Évaluation des risques de liquéfaction d’un Loess saturé Wang L.M., Yuan Z.X., Wang Q., Wu Z.J.

1647

Seismic design of retaining wall considering the dynamic response characteristic Conception sismique des murs de soutènement compte tenu des caractéristiques de réponse dynamique Watanabe K., Koseki J.

1651

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Seismic Responses of Geogrid Reinforced Wall with Tire Derived Aggregates (TDA) Backfill using Reduced-Scale Shake Table Test Les réponses sismiques des géogrilles renforcée mur avec des granulats de pneus dérivés (TDA) en utilisant emblai d’essai à échelle réduite table vibrante Xiao M., Hartman D., Ledezma M.

1655

Soil Properties of Liquefied Soils in Tokyo Bay Area by the 2011 Great East Japan Earthquake Propriétés des sols liquéfiés dans la baie de Tokyo pendant le grand séisme de l’Est du Japon en 2011 Yasuda S.

1659

In Situ Assessment of the Nonlinear Shear Modulus of Municipal Solid Waste Évaluation in situ du module non linéaire de cisaillement des déchets solides municipaux Zekkos D., Sahadewa A., Woods R., Stokoe K., Matasovic N.

1663

Analyse sismique couplée des sols renforcés par inclusions rigides Coupled dynamic analysis of soils reinforced with stiff columns Zerfa FZ.

1667

Technical committee 204 Underground Construction in Soft Ground Comité technique 204 Construction souterraine en sols mous General Report of TC204 - Underground Constructions Rapport général du TC204 - Constructions souterraines Dias T.G.S., Bezuijen A.

1673

Diametric deformations in the concrete segment lining of a tunnel excavated in soft soils. Criteria for their evaluation and mitigation actions for their control Déformations diamétrales dans le secteur du béton revêtement d’un tunnel creusé dans les sols mous. Critères de leur évaluation et des mesures d’atténuation pour leur contrôle Aguilar M.A., Valencia J.D., Schmitter J.J., Auvinet-Guichard G., Rangel-Núñez J.L. Effect of the subsoil conditions in the seismic interaction between two underground stations connected by a circular section tunnel Effet des conditions du sous-sol à l’interaction sismique entre deux stations de métro reliées par un tunnel de section circulaire Botero E., Ossa A., Ovando E., Sierra L., Giraldo V. Application of Ductile Segments to Tunnels in Close Proximity Utilisation de voussoirs ductiles à des tunnels très proches Chang J.F., Chen D.J., Moh Z.C., Yu N.T.

1681

1683

1687

Effect of pre-ground improvement method during shallow NATM tunnel excavations under unconsolidated conditions 1691 Effets de la méthode d’amélioration préalable des sols durant l’excavation de tunnel peu profond utilisant a nouvelle méthode autrichienne (NATM) dans un sol non-consolidé Cui Y., Kishida K. Field Performance of Geogrid Bridges for Stress Reduction on Buried Utilities Performance in-situ des pontages en géogrille pour réduire les contraintes dans les infrastructures souterraines El Naggar H., Turan A.

1695

Construction of a Cross Passage between Two MRT Tunnels Construction d’un passage entre deux tunnels de MRT Fang Y.S., Lin C.T., Liu C., Cheng K.H., Su C.S., Chen T.J.

1699

Auscultation et Instrumentation de démonstrateurs d’alvéoles de stockage au CMHM Monitoring and Instrumentation of demonstrators storage cells (CMHM) Gay O., Teixeira P., Bumbieler F., Morel J.

1703

Stability analyses of underground structures cut into porous limestone Contrôle de la stabilité des cavités souterraines réalisées dans le calcaire grossier Görög P., Hangodi Á., Török Á.

1707

Effect of brittle failure on deep underground excavation in eastern Taiwan Effet de la rupture fragile sur l’excavation souterraine profonde dans l’est de Taiwan Hsiao F.Y., Chi S.Y.

1711

Fast frequency-domain analysis method for longitudinal seismic response of super-long immersed tunnels Méthode d’analyse rapide dans le domaine fréquentiel pour la réponse sismique longitudinale d’un tunnel immergé à super longueur Huang M., Liu H.

1715

Field Monitoring of Shield Tunnel Lining Using Optical Fiber Bragg Grating Based Sensors Surveillance de doublure d’un tunnel au bouclier utiliser les capteurs optiques de fibre-Bragg-grating Huang A.B., Lee J.T., Wang C.C., Ho Y.T., Chuang T.S.

1719

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Contents / Table des matières

Building deformations, induced by shallow service tunnel construction and protective measures for reducing of its influence Déformations de bâtiments induites par la construction d’un tunnel de service peu profond et actions de protection pour réduire son influence Ilyichev V.A., Nikiforova N.S., Tupikov M.M.

1723

Engineering inspection and supervision of tunnels and underground stations of urban metro systems Inspection et surveillance des tunnels et stations de métro souterraines Katzenbach R., Leppla S.

1727

On the stability of a trap door evaluated by upper bound method Sur la stabilité d’une trappe évaluée par la méthode de borne supérieure Kobayashi S., Matsumoto T.

1731

Finite Element Modelling of Construction Processes of The Modular Approached Tunnelling Method Modélisation par éléments finis du processus de construction de la méthode tunnel modulared Komiya K.

1735

Cutting tool wear prognosis and management of wear-related risks for Mix-Shield TBM in soft ground Prévision d’usure des outils de coupe et management des risques liés à l’usure pour Mix-Shield TBM en terrain meuble Köppl F., Thuro K.

1739

Compensation Grouting with shallow and deep foundations – case study from the Metro B1 in Rome Injections de compensation pour les fondations superficielles et profondes – étude de cas de la ligne de métro B1 à Rome Kummerer C., Sciotti A.

1743

An evaluation of influence factors that affect pressures in backfilled trenches Une évaluation de facteurs d’influence qui affectent les pressions dans des tranchées remblayées Li L., Aubertin M., El Mkadmi N., Jahanbakhshzadeh A.

1747

Prediction of hard rock TBM penetration rate based on Data Mining techniques Modèles de prévision du taux de pénétration de tunnelier dans les roches dures Martins F.F., Miranda T.F.S.

1751

Assessment of Empirical Method Used to Study Tunnel System Performance Évaluation de la méthode empirique utilisée pour étudier la performance du système de tunnel Mazek S.A., El Ghamrawy M.K.

1755

Refurbishment and Underground Space Development of Moscow P.I. Tchaikovsky Conservatory Une reconstitution et un cosmique développement un conservatoire un Tchaïkovski moscovite souterrain Petrukhin V.P., Mozgacheva O.A., Skorikov A.V.

1759

Performance of the tunnel lining subjected to decompression effects on very soft clay deposits Performance du revêtement du tunnel soumis à des effets de décompression sur les dépôts d’argile très mous Rangel-Núñez J.L., Aguilar-Tellez M.A., Ibarra-Razo E., Paniagua W.

1763

Design of tunnel lining in consolidating soft soils Conception du revêtement des tunnels dans des sols mous en processus de consolidation Rodríguez-Rebolledo J.F., Auvinet G., Vázquez F.

1765

Effects of buried structures on the formation of underground cavity Effets des structures enterrées sur la formation d’une cavité souterraine Sato M., Kuwano R.

1769

Rational interpretation of tunneling considering existing tunnel and building loads Interprétation rationnelle du creusement des tunnels prenant en compte les tunnels préexistants et les charges iées aux constructions Shahin H.M., Nakai T., Iwata T.

1773

An elastic continuum model for interpretation of seismic behavior of buried pipes as a soil-structure interaction Un modèle de continuum élastique pour l’interprétation du comportement sismique des conduites enterrés comme une interaction sol-structure Tohda J., Yoshimura H., Maruyoshi K.

1777

Building with the Subsurface for realizing cost-efficient infrastructure Construire avec le sous-sol pour réaliser une infrastructure à coût avantageux Venmans A.A.M.

1781

Subsoil Settlement Feature of Immersed Tube Tunnel in Deep Soft Subsoil with Heavy Siltation in Open Sea Caractérisation du tassement sur sol mou de grande épaisseur d’un tunnel tube immergé soumis à un envasement important en condition de mer ouverte Xie Y., Zhang S., Zhang H., Liu B.

1785

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Case Studies of Applicability for Selection of Construction Method for Highway Underground Crossing Transit on the Deposit soils in Urban Project in Korea Étude de cas du choix de la méthode de construction pour un croisement souterrain d’autoroute sur sols meubles dans une zone urbaine en Corée Yang T.-S., Yoo N.-J., Kim S.-J., Hwang Y.-C. Numerical modeling of NATM urban tunnels and monitoring-Case study of Niayesh tunnel Modélisation numérique de tunnels urbains construits par la méthode NATM et étude de cas du contrôle du tunnel Niayesh Zolghadr E., Pasdarpour M., Majidian S., Golshani A.

1789

1793

Technical committee 205 Limit State design in Geotechnical Engineering Comité technique 205 Dimensionnement aux états limites en géotechnique General report for TC 205 - Safety and serviceability in geotechnical design: a reliability-based perspective Rapport général du TC 205 - Sécurité et maintenance en conception géotechnique : une perspective fiabiliste Salgado R.

1799

L’expérience française insolite d’un encadrement juridique : une certaine maîtrise du risque du sol Unusual French experience of a legal frame ; a certain mastery of ground risk management Carrière M.-L.

1805

Ideas for improved geotechnical structures for natural disaster mitigation Idées pour l’amélioration des ouvrages géotechniques pour l’atténuation des catastrophes naturelles Heerten G., Vollmert L.

1809

Deep Excavation in Hong Kong – Cantilever Bored Pile Wall Design Using CIRIA Report No. C580 Excavation profonde à Hong Kong Cantilever - La conception de mur paroi pieux forés Rapport Réf CIRIA Report n ° C580 Ho A., Wright M., Ng S.

1813

Comparison of the safety concepts for soil reinforcement methods using concrete columns Comparaison des concepts de sécurité pour les méthodes de renforcement de sol avec colonnes en béton Katzenbach R., Bohn C., Wehr J.

1819

Slope stability with partial safety factor method Stabilité des pentes à l’aide de la méthode de sécurité partielle Länsivaara T., Poutanen T.

1823

Assessment of embankment stability on organic soils using Eurocode 7 Évaluation de la stabilité des remblais sur sols organiques en utilisant l’Eurocode 7 Lechowicz Z., Wrzesiński G.

1827

Implementation of LRFD Methods to Quantify Value of Site Characterization Activities Mise en œuvre des méthodes de conception LRFD pour quantifier la valeur des activités de caractérisation du site Loehr J.E., Bowders J.J., Rosenblad B.L., Luna R., Maerz N., Stephenson R.W., Likos W.J., Ge L.

1831

European practice in ground anchor design related to the framework of EC7. Pratique européenne pour le dimensionnement des tirants d’ancrage en application de l’EC7 Merrifield C., Møller O., Simpson B., Farrell E.

1835

Harmonising safety and profit: ethical issues in the geotechnical activity of major consulting companies Harmoniser sécurité et profit: problèmes éthiques dans l’activité géotechnique de grosses entreprises de génie conseil Redaelli M.

1839

La norme sur les missions d’ingénierie géotechnique, clé de voûte du management des risques géotechniques de tout projet Geotechnical missions standard, the foundation of risk management for a project Robert J. Embedding Geo Risk Management. The Geo-Impuls Approach L’implantation du management des risques géotechniques. L’approche Geo-Impuls. Staveren van M.Th., Litjens P.P.T., Cools P.M.C.B.M.

1843

1847

Technical committee 206 Interactive Geotechnical design Comité technique 206 Dimensionnement géotechnique interactif General Report for TC206 - Interactive Design Rapport général du TC206 - Le dimensionnement géotechnique interactif Ho A.

XXVI

1853

Contents / Table des matières

Auscultation des fondations d’un ouvrage en terre par des capteurs à fibre optique Monitoring earthwork foundations by fibre optic sensors Artières O.

1855

A geoenvironmental application of an optimisation model Application d’un modèle d’optimisation à un problème geoenvironnemental Azimi K., Merrifield C., Gallagher E., Smith D.

1859

The role of fibre optic instrumentation in the re-use of deep foundations Rôle d’une instrumentation en fibre optique pour la réutilisation de fondations profondes Bell A., Soga K., Ouyang Y., Yan J., Wang F.

1863

Comparison of monitoring techniques for measuring deformations in an excavation Comparaison de techniques d’auscultation pour la mesure de déformations dans une excavation De Vos L., Van Alboom G., Haelterman K., Maekelberg W.

1867

Maintenance préventive des ouvrages hydrauliques par fibre optique Preventive maintenance of water retaining structures based on fiber optic systems Fry J.-J., Courivaud J.-R., Beck Y.-L., Pinettes P.

1871

Evaluation of diaphragm wall as-built data to determine the risk of leakage for the Kruisplein car park excavation in Rotterdam, The Netherlands Evaluation des données de fabrication des murs diaphragmes pour déterminer le risque de fuite dans le chantier du parking souterrain Kruisplein à Rotterdam, Pays-Bas Hannink G., Thumann V.M.

1875

Optimisation of bridge approach treatment via staged construction Optimisation du traitement de remblais d’accès à des ponts par phasage des travaux. His J.P., Carson D.J., Lee C.H.

1879

SWOT analysis Observational Method applications Analyse FFOM à l’implémentation de la méthode observationnelle Korff M., Jong de E., Bles T.J.

1883

Development of Method for Evaluating and Visualizing 3-dimensional Deformation of Earth Retaining Wall for Excavation Développement des méthodes d’évaluation et de visualisation de la déformation tridimensionnelle des murs de soutènement dans les excavations Matsumaru T., Kojima K. Geotechnical protection of engineering infrastructure objects in large cities under intense anthropogenic impact and long term operation Sécurité géotechnique d’ouvrages du génie civil sous influence anthropogène intense et exploitation à long terme Perminov N.A., Zentsov V.N., Perminov A.N. Data assimilation strategies for parameter identification of elasto-plastic geomaterials and its application to geotechnical practice Stratégie d’assimilation de données pour l’identification des paramètres de géomatériaux élastoplastiques et son applications à la pratique géotechnique Shuku T., Nishimura S., Murakami A., Fujisawa K. Experimental analyses on detection of potential risk of slope failure by monitoring of shear strain in the shallow section Analyses expérimentales sur la détection d’un risque potentiel de rupture de pente par la surveillance de la contrainte de cisaillement en pied du talus Tamate S., Hori T., Mikuni C., Suemasa N. Soutènements de grande hauteur soutenus par butons ou multi-ancrages à Monaco : de la modélisation au comportement réel Retaining wall with struts or multi-anchored for a deep excavation in Monaco: from modeling to real behaviour Utter N., Dervillé B., Beth M. New Sensing Technology and New Applications in Geotechnical Engineering Nouvelle technologie de détection et nouvelles applications à l’ingénierie géotechnique Wang Y.H., Ooi G.L., Gao Y.

1889

1893

1897

1901

1905

1909

Technical committee 207 Soil-Structure Interaction and Retaining Walls Comité technique 207 Interaction sol-structure et murs de soutènements General Report of TC 207 - Foundations and Retaining Structures Rapport général du TC 207- Fondations et ouvrages de soutènement Bilfinger W.

XXVII

1915

Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Design, Construction and Monitoring of a Mixed Soil-Reinforced and Anchored Retaining Wall in Expansive Soil Conception, construction et surveillance d’un mur mixte de sol renforcé et ancré dans un sol gonflant Abramento M., Fujii J., Cogliati B., Assakura V.

1923

Design and construction of high bermless geogrid walls in a problematic mountainous seismic region in Bulgaria Conception et construction de murs renforcés par des géogrilles de grande hauteur et sans risberme dans une région montagneuse sismique problématique en Bulgarie Alexiew D., Hangen H.

1927

La fouille de la tour Odéon à Monaco : les quatre éléments remarquables de sa conception The Deep Excavation of the Odéon Tower in Monaco: The four outstanding elements in its design Baghery S.

1931

3D Finite Element Analyses for a Laterally Loaded Pile Wall in Marine Environment– Case History Analyses 3D par éléments finis pour un mur de quai chargés latéralement dans un port – Etude de cas Bahr M.A., Tarek M.F., El-Ghamrawy M.K., Abouzaid K.S., Shaarawi E.M.

1935

Design and construction of a coffer dam on Narmada River for Indira Sagar project in central India: a case study of innovative foundation Conception innovante et construction d’un batardeau provisoire pour le barrage sur la rivière Narmada dans le cadre du projet Sagara en Inde centrale Bidasaria M.

1939

Anchored sheet pile wall design in expansive soils Conception d’un mur de palplanches ancré dans les sols expansifs Bilgin Ö., Mansour E.

1943

Performance of Soil Nails in Weathered Granite and Fill Performance de renforcement par clouage du granite altéré et du remblai Chow C.-M., Chee-Meng, Tan Y.-C.

1947

Effects on adjacent buildings from diaphragm wall installation Effets sur des bâtiments adjacents liés à l’installation de parois moulées Comodromos E.M., Papadopoulou M.C., Konstantinidis G.K.

1951

Prise en compte des effets de bord dans un massif renforcé par inclusions rigides Modeling edge effects at the periphery of a rigid inclusion reinforced soil volume Cuira F., Simon B.

1955

Influence of facing vertical stiffness on reinforced soil wall design Influence de la rigidité verticale du parement dans la conception des murs en sols renforcés Damians I.P., Lloret A., Josa A., Bathurst R.J.

1959

Earth Pressure from Strip Footings on an Anchored Sheet Pile Wall Poussée des terres provenant de semelles filantes sur un mur de palplanches ancré Denver H., Kellezi L.

1963

Top Down Construction Alongside Of Bosphorus - A Case Study Construction en « Top - Down » le long du Bosphore - Une étude de cas Durgunoglu T., Kulac F., Ikiz S., Akcakal O.

1967

Experiences with SBMA ground anchors in spanish soils Etude expérimentale avec les tirants d’ancrage SBMA dans le sol espagnol Fernandez Vincent J.M.

1971

Computed and observed ground movements during top-down construction in Chicago Mouvements de terrains calculés et observés en construction descendante à Chicago Finno R.J., Arboleda L., Kern K., Kim T., Sarabia F.

1975

Comparative Life Cycle Assessment of Geosynthetics versus Concrete Retaining Wall Analyse de cycle de vie comparative d’un épaulement géotextile et conventionnel Frischknecht R., Büsser-Knöpfel S., Itten R., Stucki M., Wallbaum H.

1979

Design of inverted T-shape Cantilever Wall a Relief Floor Concption d’un mur équerre avec dalle de délestage Ganne P.P., Raucroix X.

1983

An Anchored Retaining Wall in CSM Un soutènement ancré en CSM Gomes Correia A., Tinoco J., Pinto A., Tomásio R.

1987

Conception, modélisation et auscultation d’une très grande excavation à Monaco Design, modelization and monitoring for a very large excavation in Monaco Guilloux A., Porquet M., De Lavernée P., Lyonnet P., Roman P.

1991

A Case Study of 3D FE Analysis of a Deep Excavation Adjacent to a Tunnel Construction Une étude de cas d’une simulation tridimensionnelle d’analyse par éléments finis d’une excavation profonde adjacente à une construction d’un tunnel Guler E., Osmanoglu U., Koç M.

1995

XXVIII

Contents / Table des matières

Suction Caisson Installation in Shallow Water: Model Tests and Prediction Installation de caissons à succion en eau peu profonde: essais et prédiction Guo W., Chu J.

1999

Instrumentation de la paroi moulée du bassin de Blanc-Mesnil : retro-analyse et calage des modèles de calcul Instrumentation of the diaphragm wall of the Blanc-Mesnil Basin : retro-analysis and calibration of calculation models Gutjahr I., Doucerain M., Schmitt P., Heumez S., Maurel C.

2003

Displacement of an apartment building next to a deep excavation in Rotterdam Déplacements d´un bâtiment d’habitation adjacent à un chantier profond d’excavation à Rotterdam Hannink G., Oung O.

2007

Calculation method of optimization the soil-cement mass dimensions to reduce the enclosure displacements in deep excavation Calcul des dimensions optimales du massif du sol-ciment pour réduire les déplacements de fouilles profondes Ilyichev V.A., Gotman Y.A.

2011

Case Studies of Complicate Urban Excavation from Design to Construction Études de cas d’excavations complexes en site urbain: de la conception à la construction Jang Y.S., Choi H.C., Shin S.M., Kim D.Y.

2015

Passive Pressure on Skewed Bridge Abutments Pression passive sur des culées de pont asymétriques Jessee S., Rollins K.

2019

Deformation behaviour of clay due to unloading and the consequences on construction projects in inner cities Étude du comportement en déformations de l’argile suite à un retrait de charge et conséquences lors de projets de constructions en zone urbaine Katzenbach R., Leppla S.

2023

Large tailings heaps and the influence on infrastructures due to the resulting soil deformation Les grands terrils miniers et leur influence sur les infrastructures voisines à travers la déformation des sols Katzenbach R., Leppla S., Seip M., Schleinig J.-P., Schnürer F.

2027

In-situ tests of permanent prestressed ground anchors with alternative designs of anchor bond length Essais in situ des tirants d’ancrage précontraints permanents avec des conceptions alternatives de la longueur de scellement Klemenc I., Logar J.

2031

Response of piled buildings to deep excavations in soft soils Déformations des bâtiments liés aux excavations profondes situé dans les sols mous Korff M., Mair R.J.

2035

Deep excavation in Irish glacial deposits Excavation profonde des dépôts glaciaires Irlandais Long M., O’Leary F., Ryan M., Looby M.

2039

Active earth thrust on walls supporting granular soils: effect of wall movement Pression active des terres sur des murs soutenant des sols granulaires: l’effet du mouvement du mur Loukidis D., Salgado R.

2043

Innovative solutions for supporting excavations in slopes Solutions innovantes pour le soutien d’excavations situées dans des terrains en pente Lüftenegger R., Schweiger H.F., Marte R.

2047

Design and Construction of Inclined-Braceless Excavation Support Applicable to Deep Excavation Dimensionnement et construction du support d’excavation Incliné sans butons applicable à une excavation profonde Maeda T., Shimada Y., Takahashi S., Sakahira Y.

2051

Shaking table tests on caisson-type quay wall with stabilized mound Essais à table vibrante sur les murs de quai de type caisson avec butte stabilisée Mizutani T., Kikuchi Y.

2055

Inspection of structural health of existing railway retaining walls Inspection de l’état structurel des murs de soutènement des voies de chemin de fer existantes Nakajima S., Shinoda M., Abe K.

2059

Mechanism of Settlement Influence Zone due to Deep Excavation in Soft Clay Mécanisme de la zone d’influence de tassement dû à une excavation profonde dans l’argile molle Ou C.-Y., Teng F.-C., Hsieh P.-G., Chien S.-C.

2063

Establishing a high risk construction pit in a hurry L’établissement d’une excavation profonde à risque élevé en court temps Philipsen J.

2067

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Innovative Solution of King Post Walls combined with CSM Panels Solution Innovante de Parois Berlinoise combinée avec des Panneaux de CSM Pinto A., Tomásio R., Godinho P.

2071

Unusual Geotechnical Solutions at the Leixões Cruise Terminal Solutions géotechniques inhabituelles au terminal de croisières de Leixões Pinto A., Pita X., Neves M., Vaz J.

2075

Aspects on designing and monitoring a deep excavation for a highly important structure Aspects de conception et de suivi d’une excavation profonde d’une très importante structure Popa H., Manea S., Batali L., Olteanu A.

2079

FEM-aided design of a novel device for soil anchoring Conception assistée par éléments finis d’un nouveau système pour l’ancrage des sols Prisco di C., Pisanò F.

2083

Structural Optimization in Geotechnical Engineering Optimisation de la structure dans la géotechnique Pucker T., Grabe J.

2087

Role of the facing on the behaviour of soil-nailed slopes under surcharge loading Rôle du parement sur le comportement des pentes de sol cloué sous surcharge Sanvitale N., Simonini P., Bisson A., Cola S.

2091

Geotechnical aspects in sustainable protection of cultural and historical monuments Les aspects géotechniques dans le développement durable des monuments historiques et culturels Sesov V., Cvetanovska J., Edip K.

2095

Various use of diaphragm walls for construction of multilevel road junction – Design and monitoring of displacements Diverses utilisations de parois moulées pour la construction de l’intersection des routes à plusieurs niveaux – Conception et le suivi des déplacements Siemińska-Lewandowska A., Mitew-Czajewska M., Tomczak U.

2099

Effects of plane shapes of a cofferdam on 3D seepage failure stability and axisymmetric approximation Effets des formes planes d’un batardeau sur la stabilité après une rupture par infiltration tridimensionnelle et sur l’approximation axisymétrique Tanaka T., Kusumi S., Inoue K.

2103

Stability and dewatering problems of deep excavations in Bratislava Les problèmes de stabilité et d’assèchement des excavations profondes dans la ville de Bratislava Turček P., Frankovská J., Súľovská M.

2107

Managed remediation of a large Victorian gravity quay wall using the observational method Stabilisation d’un grand mur de quai de l’époque Victorienne gérée en utilisant la méthode observationnelle Turner M. J, Smith N A.

2111

Concrete panel walls – Current development on interaction of earthworks, geosynthetic reinforcement and facing Comportement des parements béton de murs de soutènement en sols renforcés – Interaction entre les sols remblayés, le renforcement et le parement Vollmert L., Niehues C., Pachomow D., Herold A., Verstraaten W.

2115

The influence of bond stress distribution on ground anchor fixed length design. Field trial results and proposal for design methodology L’influence de la répartition des contraintes sur les tirants d’ancrage de longueur fixe. Résultats de planche d’essais et proposition de méthodologie de conception Vukotić G., González Galindo J., Soriano A.

2119

The sustainability and assessment of drystone retaining walls Le développement durable et l’évaluation des murs de soutènement en pierres sèches Warren L., McCombie P., Donohue S.

2123

Numerical modelling of groundwater flow around contiguous pile retaining walls Modélisation numérique des écoulements des eaux souterraines autour d’écrans de soutènement de pieux contigusë Wiggan C.A., Richards D.J., Powrie W.

2127

Geosynthetic Reinforced Soil Wall Performance under Heavy Rainfall La performance du mur en sol renforcé par géosynthétiques sous de fortes pluies Yoo C., Jang D.W.

2131

Technical committee 208 Slope Stability in Engineering Practice Comité technique 208 Stablité des pentes pour la pratique de l’ingéneiur General Report of TC 208 - Slope Stability in Engineering Practice Rapport Général du TC 208 - La stabilité des talus dans la pratique de l’ingénieur Bowman E.T., Fannin R.J.

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Contents / Table des matières

Calculation of slopes stability based on the energy approach Calcul de la stabilité des pentes sur la base de l’approche énergetique Bogomolov A.N., Ushakov A.N., Bogomolova O.A.

2145

Preservation of slope stability along the by-pass Vlora Maintien de la stabilité des pentes dans le contournement de « vlora » Bozo L., Allkja S.

2149

A Methodology for Evaluating Liquefaction Susceptibility in Shallow Sandy Slopes Une méthodologie pour l’évaluation de susceptibilité à la liquéfaction dans les pentes sableuses Buscarnera G., Whittle A.J.

2153

Assessment of landslide run-out by Monte Carlo simulations Évaluation de la dynamique des glissements de terrain par des simulations de Monte-Carlo Cepeda J., Quan Luna B., Nadim F.

2157

The Challenge of the Slope Failure Problem and Its Remedial Considerations at Mileage 39km, Mt. Ali Road, Taiwan Le défi du problème du glissement de pente et des dispositions correctives apportées, au kilomètre 39, de la route Mt. Ali Road, àTaïwan Chang M., Huang R. Downstream Frontal Velocity Reduction Resulting from Baffles Effets des déflecteurs dans la réduction des vitesses frontales dans un écoulement descendant. Choi C.E., Ng C.W.W., Kwan J.S.H., Shiu H.Y.K., Ho K.K.S., Koo R.C.H. GPS instrumentation and remote sensing study of slow moving landslides in the eastern San Francisco Bay hills, California, USA Instrumentation GPS et télédétection de glissements de terrains lents dans les collines est de la baie de San Francisco, Californie, USA Cohen-Waeber J., Sitar N., Bürgmann R. Geotechnical Characterization, Stability Analysis, and the Stabilization Process for a Landslid in a area of Barreiras Formation and Granite Residual Soils, Pernambuco Caractérisation géotechnique, analyse de la stabilité et procédés de stabilisation pour un glissement de terrain dans des matériaux du type « Barreiras Formation » et sols de granite résiduel, Pernambuco Coutinho R.Q., Silva da M.M.

2161

2165

2169

2173

Progressive failure of slopes with sensitive clay layers Rupture progressive de pentes comportant des couches d’argile sensible Dey R., Hawlader B., Phillips R., Soga K.

2177

Quantitative vulnerability estimation for individual landslides Estimation quantitative de la vulnérabilité aux glissements de terrain Du J., Yin K., Nadim F., Lacasse S.

2181

A site specific early warning system for rainfall induced landslides Utilisation d’un site spécifique pour l’élaboration d’un système d’alerte rapide pour les instabilités de pente induites par les pluies. Harris S., Orense R., Itoh K.

2185

Characteristics of Ground Motion on Colluviums Slope Induced by Heavy Rainfall Caractéristiques du déplacement du sol sur la pente de colluvions induit par la pluie violente Jeng C.J., Sue D.Z.

2189

Stability and movements of open-pit lignite mines in Northern Greece Stabilité et mouvements de terrain dans les mines de lignite à ciel ouvert en Grèce du Nord Kavvadas M., Agioutantis Z., Schilizzi P., Steiakakis C.

2193

A web-based tool for ranking landslide mitigation measures Un outil internet pour classer les techniques visant à diminuer le risque de glissements de terrain Lacasse S., Kalsnes B., Vaciago G., Choi Y.J., Lam A.

2197

A Numerical Study of Granular Surge Flow through a Row of Baffles Une étude numérique des écoulements granulaires à travers une rangée de chicanes Law R.P.H., Lam A.Y.T., Choi K.Y.

2201

Full-Scale Field Monitoring of a Rainfall-Induced Sliding Slope in Hainan, China Étude en vraie grandeur d’un talus glissant soumis à des précipitations à Hainan en Chine Li A.G., Qiu J.J., Mo J.F., Gao W., Tham L.G., Yan R.W.M.

2205

Estimation and Prediction of Debris Flow Potential Using Discrimination Analysis Estimation et prédiction du potentiel d’écoulement de boue utilisant une analyse discriminante Lin M.L., Lin Y.S.

2209

Value of Landslide Investigation to Geotechnical Engineering Practice in Hong Kong Ingénierie des glissements de terrain à Hong Kong Lo D.O.K., Lam H.W.K.

2213

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Analyses of Seismic Slope Stability and Subsequent Debris Flow Modeling Analysis de stabilité de pente sous sollicitation sismique et modélisation des écoulements de boues induits Long X., Tjok K.-M.

2217

Quantitative risk assessment for earthquake-triggered landslides using Bayesian network Évaluation quantitative du risque associé aux glissements de terrain déclenchés par séisme en utilisant un réseau Bayésien Nadim F., Liu Z.Q.

2221

Collapse behavior of slope due to change in pore water pressure Effondrement d’une pente à cause d’une variation de la pression interstitielle Nakata Y., Kajiwara T., Yoshimoto N.

2225

Future evolution of slope stability analysis created by SPH method Évolution future de l’analyse de stabilité des pentes créé par la méthode SPH Nonoyama H., Yashima A., Moriguchi S.

2229

Slope stability along a new road “Drisht –Drisht castle” Stabilité de pente le long de la nouvelle route «Drisht-Drisht castle» Paçi E., Cullufi H., Dervishaj A.

2233

Landslides induced by the interaction of an earthquake and subsequent rainfall. A spatial and temporal model Glissements de terrain induits par l’interaction d’un tremblement de terre suivi de précipitations – Un modèle spatio-temporel. Quan Luna B., Vangelsten B.V., Liu Z.Q., Eidsvig U., Nadim F.

2237

Analyse des glissements de terrains en région urbanisée Analysis of landslides in urban regions Rahhal M.E., Hajal M., Seoud J.P., Rafie E.

2241

A smart adaptive multivariable search algorithm applied to slope stability in locating the global optima Un algorithme adaptatif multivariable de recherche d’optimum global appliqué à la stabilité des pentes Saha A.

2245

Soil slope stability of hydropower reservoirs - from geological site investigation to design of mitigation measures La stabilité des talus de réservoirs hydroélectriques - de l’investigation géologique du site à la conception de mesures d’atténuation Saurer E., Prager C., Marcher T.

2249

Landslide stabilization by piles: A case history Stabilisation des glissements de terrain par des pieux: un cas d’étude Şengör M.Y., Ergun M.U., Huvaj N.

2253

Landslide Susceptibility Mapping Using Bayesian Conditional Probability Model at An Linh Commune, Tuy An District, Phu Yen Province, Vietnam Élaboration de la carte de risques de glissement de terrain sur la commune de An Linh, district de Tuy An, province de Phu Yen, à l’aide d’un modèle Bayesien de probabilité conditionnelle Son N.T., Ha P.T.S., Son L.M.

2257

Influence of Ground Motion Variability on Seismic Displacement Uncertainty Influence de la variabilité des mouvements de terrain sur l’incertitude des déplacements en régime sismique Strenk P.M., Wartman J.

2261

A new approach to assess the potential for flow slide in sensitive clays Une nouvelle approche pour évaluer le potentiel d’écoulement des argiles sensibles Thakur V., Degago S.A., Oset F., Dolva B.K., Aabøe R.

2265

Landslide risk assessment in the Göta river valley: effect of climate changes L’évaluation des risques de glissement de terrain dans la vallée de la rivière Göta : effet des changements climatiques Tremblay M., Svahn V., Lundström K.

2269

Deformation and water seepage observed in a natural slope during failure process by artificial heavy rainfall Déformation du sol et infiltration d’eau observes le long d’une pente naturelle pendant le processus de glissement dû à de fortes pluies artificielles Uchimura T., Gizachew G., Wang L., Nishie S., Seko I.

2273

Study on field detection and monitoring of slope instability by measuring tilting motion on the slope surface Détection et surveillance in situ des phénomènes d’instabilités de pente par mesure locale des mouvements de surface Wang L., Nishie S., Seko I., Uchimura T.

2277

The physical vulnerability of roads to debris flow La vulnérabilité physique des routes aux coulées de boue Winter M.G., Smith J.T., Fotopoulou S., Pitilakis K., Mavrouli O., Corominas J., Agyroudis S.

2281

Inspection and Capacity Assessment of Anchored Slopes Inspection et évaluation des capacités des pentes renforcées par ancrage Yeh H.n-S., Wang C.-S., Wei C.-Y., Lee S.-M., Ho T.-Y., Hsiao C.-A., Tsai L.-S.

2285

XXXII

Contents / Table des matières

2011 Seoul Debris Flow and Risk Analysis Coulée de boue à Séoul en 2011 et analyse des risques Yune C.-Y., Kim G., Lee S.W., Paik J.

2289

Technical committee 209 Offshore Geotechnics Comité technique 209 Géotechnique marine General Report of TC209 - Offshore Geotechnics Rapport général du TC209 - Géotechnique Offshore Jewell R.A.

2295

Shallow foundations for offshore wind towers Fondations superficielles pour des installations éoliennes maritimes Arroyo M., Abadías D., Alcoverrro J., Gens A.

2303

Modelling of monopile-footing foundation system for offshore structures in cohesionless soils Modélisation d’un système de fondation superficielle isolé pour sur les structures maritimes dans les sols pulvérulents Arshi H.S., Stone K.J.L., Vaziri M., Newson T.A., El-Marassi M., Taylor R.N., Goodey R.J.

2307

Influence of jack-up footprints on mudmat stability – How beneficial are 3D effects? Influence des dépressions laissées par les jack-ups sur la capacité portante des mudmats – quels sont les effets bénéfiques d’une analyse en 3D? Ballard J.-C., Charue N.

2311

Design and installation of buried large diameter HDPE pipelines in a coastal area Project et installation de tuyaux enterrés de grand diamètre en zone côtière Bellezza I., Mazzier F., Pasqualini E., D’Alberto D., Caccavo C., Serrani C.

2315

Geotechnical Exploration for Wind Energy Projects Compagnes géotechniques destinées aux parcs éoliens Ben-Hassine J., Griffiths D.V.

2319

Essais cycliques axiaux sur des pieux forés dans des sables denses Cyclic axial load tests on bored piles in dense sands Benzaria O., Puech A., Le Kouby A.

2323

Essais cycliques axiaux sur des pieux forés dans l’argile surconsolidée des Flandres Cyclic axial load tests on bored piles in overconsolidated Flanders clay Benzaria O., Puech A., Le Kouby A.

2327

Fondations superficielles glissantes pour l’offshore profond – Méthodologie de dimensionnement Deep Offshore Sliding Footings – Design Methodology Bretelle S., Wallerand R.

2331

Proposition d’une loi t-z cyclique au moyen d’expérimentations en centrifugeuse Proposal of cyclic t-z law by means of centrifuge experiments Burlon S., Thorel L., Mroueh H.

2335

Deformation behavior of single pile in silt under long-term cyclic axial loading Comportement d’un pieu isolé sous chargement axial cyclique de longue durée dans un limon Chen R.P., Ren Y., Zhu B., Chen Y.M.

2339

Time-Varying Dynamic Properties of Offshore Wind Turbines Evaluated by Modal Testing Étude expérimentale de l’évolution temporelle des propriétés dynamiques d’éoliennes maritimes Damgaard M., Andersen J.K.F., Ibsen L.B., Andersen L.V.

2343

Numerical investigation of dynamic embedment of offshore pipelines Étude numérique de l’ancrage dynamique de conduites enterrées maritimes Dutta S., Hawlader B., Phillips R.

2347

Post Cyclic Behaviour of Singapore Marine Clay Le comportement post-cyclique de l’argile marine de Singapour Ho J., Goh S.H., Lee F.H.

2351

Centrifuge test and numerical modeling for a suction bucket monopod foundation Essai en centrifugeuse et la modélisation numérique d’une fondation de type : caisson à succion Kim D.J., Youn J.U., Jee S.H., Choi J., Choo Y.W., Kim S., Kim J.H., Kim D.S., Lee J.S.

2355

A large deformation finite element analysis solution for modelling dense sand Solution d’analyse par éléments finis d’une large déformation pour la modélisation de sable dense Li X., Hu Y, White D.

2359

Plugging Effect of Open-Ended Displacement Piles Prise en compte de l’effet de bouchon pour les pieux battus ouverts Lüking J., Kempfert H.-G.

2363

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

A simplified procedure to assess the dynamic stability of a caisson breakwater Une procédure simplifiée pour évaluer la stabilité dynamique d’une digue en caissons Madrid R., Gens A., Alonso E., Tarrago D.

2367

The new remediation technique for buried pipelines under permanent ground deformation Une nouvelle technique de pose des conduites enterrées soumises à des déformations permanentes du sol Moradi M., Galandarzadeh A., Rojhani M.

2371

Site investigation and geotechnical design strategy for offshore wind development Investigation géotechniques et stratégie de conception pour le développement d’éoliennes maritimes Muir Wood A., Knight P.

2375

Diagrammes de stabilité cyclique de pieux dans les sables Cyclic stability diagrams for piles in sands Puech A., Benzaria O., Thorel L., Garnier J., Foray P., Silva M., Jardine R.

2379

Utilisation des essais d’expansion cyclique pour définir des modules élastiques en petites déformations Determining small strain elastic modulus using cyclic expansion tests Reiffsteck P., Fanelli S., Tacita J.-L., Dupla J.-C., Desanneaux G.

2383

Displacement response to axial cyclic loading of driven piles in sand Réponse en déplacement au chargement cyclique axial de pieux battus dans le sable Rimoy S., Jardine R., Standing J.

2387

Experimental Testing of Monopiles in Sand Subjected to One-Way Long-Term Cyclic Lateral Loading Étude expérimentale de monopiles dans le sable soumis à un chargement cyclique transversal non alterné Roesen H.R., Ibsen L.B., Andersen L.V.

2391

Pieu sous charge latérale : développement de lois de dégradation pour prendre en compte l’effet des cycles Pile cyclic lateral loading: Development of degradation laws for describing the cyclic effect Rosquoët F., Thorel L., Garnier J., Chenaf N.

2395

Behavior of marine silty sand subjected to long term cyclic loading Comportement du sable limoneux marin soumis à une charge cyclique de longue durée Safdar M., Kim J.M.

2499

Influence des chargements cycliques axiaux dans le comportement et la réponse de pieux battus dans le sable Influence of cyclic axial loads in the behaviour and response of driven piles in sand Silva M., Foray P., Rimoy S., Jardine R., Tsuha C., Yang Z.

2403

Characterization of the geotechnical properties of a carbonate clayey silt till for a shallow wind turbine foundation Caractérisation des propriétés géotechniques d’un silt argileux carbonaté glaciaires pour une fondation superficielle d’éolienne Tyldesley M., Newson T., Boone S., Carriveau R.

2407

Cyclic loading of caisson supported offshore wind structures in sand Chargement cyclique des éoliennes offshore soutenues par des caissons à succion en sable Versteele H., Stuyts B., Cathie D., Charlier R.

2411

Technical committee 211 Ground Improvement Comité technique 211 Amélioration des sols General Report of TC 211 - Ground Improvement Rapport général du TC 211 - Amélioration des sols Huybrechts N., Denies N.

2417

Time-dependent behaviour of foundations lying on an improved ground Temps-comportement dépendant de fondations reposant sur un sol amélioré Alupoae D., Aşuencei V., Răileanu P.

2425

Centrifugal and numerical analysis of geosynthetic-reinforced soil embankments Étude par centrifugeuse et analyse numérique des remblais renforcés par géotextile Bo L., Linli J., Ningyu Z., Sinong L.

2429

Compacted soil columns for foundations on collapsible soils. Laboratory and in-situ experimental study Colonnes de sols compactés utilisées pour des fondations sur sols effondrables. Étude expérimentale menée en laboratoire et in-situ Burlacu C., Olinic E., Manea S., Uţă P

2433

Selected problems connected with the use of the jet grouting technique Certains problèmes liés à l’application de la technologie d’injection de jet Bzówka J., Juzwa A., Wanik L.

2437

Column Supported Embankments for Transportation Infrastructures: Influence of Column Stiffness, Consolidation Effects and Cyclic Loading

XXXIV

2441

Contents / Table des matières

Remblais sur sols renforcés avec de colonnes ballastées pour les infrastructures de transport: Influence de la rigidité des colonnes, des effets de consolidation et du chargement cyclique Carvajal E., Vukotić G., Sagaseta C., Wehr W. Foundations of embankments using encased stone columns Fondations de remblais avec des colonnes ballastées entourées de géotextile Castro J., Sagaseta C., Cañizal J., Da Costa A., Miranda M.

2445

Consolidation theory for combined vacuum pressure and surcharge loading Théorie de la consolidation sous l’action combinée du vide et d’un pré-chargement Chai J.-C., Carter J. P.

2449

Displacement rigid inclusions Inclusions rigides refoulées Cirión A., Paulín J., Racinais J., Glandy M.

2453

Prediction of the unconfined compressive strength in soft soil chemically stabilized Prévision de la résistance à la compression non confinée dans sols mous chimiquement stabilisés Correia A.A.S., Venda Oliveira P.J., Lemos L.J.L.

2457

Modélisation numérique du comportement d’une colonne de soil-mixing et confrontation à un essai de chargement en vraie grandeur Numerical modeling of a soil-mixing column behavior and comparison with a full-size load test Cuira F., Costa d’Aguiar S., Grzyb A., Pellet F., Mosser J.-F., Guimond-Barrett A., Le Kouby A.

2461

Design of Deep Soil Mix Structures: considerations on the UCS characteristic value Dimensionnement des structures en soil mix : considérations sur la valeur caractéristique UCS Denies N., Van Lysebetten G., Huybrechts N., De Cock F., Lameire B., Maertens J., Vervoort A.

2465

Method of improvement of the subsoil under Adora facility – Ohrid, Republic Of Macedonia Méthode d’amélioration du sous-sol sous le bâtiment Adora – Ohrid, République de Macédoine Dimitrievski L., Ilievski D., Dimitrievski D., Bogoevski B., Strasheski A.

2469

Geoencased columns: toward a displacement based design Colonnes renforcée par géotextiles: vers une conception basée sur le déplacement Galli A., Prisco di C.

2473

Design prediction of the strengthened foundation base deformation by field tests data La prèvision de calcul des déformations de la base des fondements reportès à partir des recherches prises en nature Gotman A., Gotman N.

2477

Standardization of the molding procedures for stabilized soil specimens as used for QC/QA in Deep Mixing application Normalisation des procédures pour la production d’éprouvettes de sols stabilisés utilisées dans les processus de QC/QA pour des applications de « Deep Mixing » Grisolia M., Leder E., Marzano I.P.

2481

Analysis of Floating Pile Capacity in Improved Ground for Thi Vai Port, Vietnam Analyse de la capacité de Pile flottant dans un terrain Thi Vai Amélioration de Port, Vietnam Hai N.M., Tuong N.K., Long P.D., Nhon P.V.

2485

Carbonate Cementation via Plant Derived Urease Cimentation carbonatée par l’utilisation d’uréase issue de plantes Hamdan N., Kavazanjian Jr. E., O’Donnell S.

2489

Experimental investigation on bearing capacity of geosynthetic encapsulated stone columns Étude expérimentale sur la capacité portante des colonnes de pierre géosynthétiques encapsulées Hataf N., Nabipour N.

2493

Performance and Prediction of Vacuum Consolidation Behavior at Port of Brisbane Avantages et prédictions de comportement due a la consolidation sous vide au port de Brisbane Indraratna B., Rujikiatkamjorn C., Geng X., Ameratunga J.

2497

Improvement of a Clay Deposit using Prefabricated Vertical Drains and Pre-loading - A Case Study Amélioration d’un massif d’argile à l’aide de drains verticaux préfabriqués et de pré-chargement Une étude de cas Islam M.S., Yasin S.J.M.

2501

Importance et applications des inclusions de grande inertie Importance and practical examples of inertial soil improvement. Jeanty J.M., Mathieu F., Benhamou L., Berthelot P.

2505

Assessement of Carillo’s theory for improved Tunis Soft Soil by Geodrains (manque traduction en français) Jebali H., Frikha W., Bouassida M.

2509

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Improvement of soft fat clay using rigid inclusions and vertical drains Amélioration d’une argile plastique molle par inclusions rigides et drains verticaux Kirstein J.F., Wittorf N.

2513

Interaction of stone column and surrounding soil during its construction: 3D numerical analysis Interaction d’une colonne ballastée et du sol environnant pendant sa construction : analyse numérique 3D Klimis N.S., Sarigiannis D.D.

2517

Laboratory tests and numerical modeling for embankment foundation on soft chalky silt using deep-mixing Essais au laboratoire et modélisation numérique de la fondation d’un remblai sur un limon crayeux mou des sols améliorés par malaxage en profondeur Koch E., Szepesházi R.

2521

Assessment of bio-mechanical reinforcement materials influencing slope stability, based on numerical analyses Évaluation des matériaux de renforcement bio-mécaniques qui influencent la stabilité des pentes par des analyses numériques Koda E., Osinski P.

2525

Evaluation of Vertical Drain-enhanced Radial Consolidation with Modified Analytical Solution Évaluation de la consolidation radiale améliorée par des drains verticaux par une solution analytique modifiée Lee C., Choi Y., Lee W., Hong S.J.

2529

Adjusting the soil stiffness with stabilisation to minimize vibration at Maxlab IV – Asynchrotron radiation facility in Sweden Ajustement de la rigidité du sol par stabilisation pour minimiser les vibrations à Maxlab IV, un centre de rayonnement synchrotron en Suède Lindh P., Rydén N. Construction and Performance of Containment Bund Using Geotextile Tubes Filled With Cement Mixed Soil in Singapore La construction et la performance de la digue de confinement utilisant des tubes géotextiles remplis de terre mélangée au ciment à Singapour Loh C.K., Chew S.H., Tan C.Y., Lim S.K., Lam J.P.W. Reinforcement of completely decomposed granite with discrete fibres Renforcement de granite complètement décomposé avec des morceaux fibres Madhusudhan B.N., Baudet B.A. Hybrid Application of Deep Mixing Columns Combined with Walls as a Soft Ground - Improvement Method Under Embankments Application hybride de la méthode de « Deep Mixing » sur des colonnes combinées à des murs en tant que méthode d’amélioration des sols mous sous remblais Matsui H., Ishii H., Horikoshi K.

2533

2537

2541

2545

Application of cement deep mixing method for underpinning Application de colonnes de sol-ciment pour travaux de reprise en sous œuvre Melentijevic S., Arcos J.L., Oteo C.

2549

Lime Remediation of Reactivated Landslides Traitement à la chaux pour la stabilisation des glissements réactivés Mesri G., Moridzadeh M.

2553

Improvement of the Soil under the Concrete Pavement of a Plant’s Hall Amélioration du terrain d’assise sous la dalle en béton d’une halle d’usine Mihova L., Kolev Ch.

2557

Effect of Smear on Strength Behavior of SCP-Reinforced Soft Ground Effet de comportement de l’étalement de force du SCP- Sol mou renforcé Mir B.A., Juneja A.

2561

Bio-mediated soil improvement utilized to strengthen coastal deposits Amélioration du sol biologiquement négociée utilisée pour renforcer les dépôts côtiers Montoya B.M., Feng K., Shanahan C.

2565

Effect of Grout Bleed Capacity on the Engineering Properties of Cement Grouted Sands Effet de la capacité de ressuage de coulis de ciment sur les propriétés mécaniques des sables injectés Pantazopoulos I.A., Atmatzidis D.K., Basas V.G., Papageorgopoulou S.K.

2569

Numerical Analysis to Quantify the Influence of Smear Zone Characteristics on Preloading Design in Soft Clay Analyses numériques pour quantifier l’influence des caractéristiques  de la zone endommagée sur la conception de préchargement dans les argiles molles Parsa-Pajouh A., Fatahi H., Khabbaz B.

2573

Construction of virtual sites for reliability-based design Construction de sites virtuels à des fins de conception fiabiliste Phoon K.K., Ching J.

2577

XXXVI

Contents / Table des matières

Technique of reinforced soil base calculation under fall initiation in ground mass Technique du compte armé les raisons du sol à l’apparition des échecs à le massif du sol Ponomaryov A., Zolotozubov D.

2581

Stress Concentration Ratio and Design Method for Stone Columns using 2D FEA with Equivalent Strips Ratio de concentration de contraintes et méthode de conception pour les colonnes ballastées en utilisant une analyse aux éléments finis 2D avec des bandes équivalentes Poon B., Chan K.

2585

Porosity/cement index to evaluate geomechanical properties of an artificial cemented soil Le paramètre porosité/ciment pour l’évaluation des propriétés géomécaniques d’un sol cimenté artificiellement Rios S., Viana da Fonseca A.

2589

Compressive Strength of Fiber-Reinforced Lightly-Cement Stabilized Sand Résistance à la compression des sables renforcées par fibres et ciment Sadek S., Najjar S., Abboud A.

2593

Conservatoriumhotel Amsterdam, geotechnical design and monitoring Conservatoriumhotel Amsterdam, conception géotechnique et instrumentation Stoel van der A.E.C., Vink D., Bouma J.

2597

Impact of the soil-stabilization with lime Impact de la stabilisation des sols à la chaux Szendefy J.

2601

Etude paramétrique en laboratoire du matériau Deep Soil-Mixing Laboratory parametric study of the Deep Mixing material Szymkiewicz F., Le Kouby A., Reiffsteck P., Mosadegh A., Tacita J.-L.

2605

Investigation of failure analysis of clay reinforced with sand encapsulated Enquête sur l’analyse des défaillances d’argile renforcé avec du sable enrobe Tabarsa A.R., Hajiesmaeilian S.

2609

Influence of relative density on microbial carbonate precipitation and mechanical properties of sand L’influence que la densité relative du sol donne dans précipitation du carbonate microbienne et propriétés de la mécanique Tsukamoto M., Inagaki T., Sasaki Y., Oda K.

2613

The reinforcement of soils by dispersed oversized particles Le renforcement des sols par les particules trop grandes non réparties uniformément Vallejo L.E., Lobo-Guerrero S., Seminsky L.F., Caicedo B.

2617

Analysis of Displacements of GPA in Normally Consolidated Soft Soil L’analyse des déplacements des GPA dans le sol mou normalement consolidé Vidyaranya B., Madhav M.R.

2621

Bridge foundation on very soft alluvia with stone column ground improvement Fondation de pont sur alluvions très mous et amélioration du sol avec des colonnes ballastées Vlavianos G.J., Marinelli A.K., Andrianopoulos K., Foti S.

2625

Subgrade improvement measures for the main rescue roads in the urban redevelopment area HafenCity in Hamburg Mesures d’amélioration du sol de fondation des principales routes de secours dans la zone du réaménagement urbain de la HafenCity à Hamburg Weihrauch S., Oehrlein S., Vollmert L.

2629

Fiber Reinforced Cement Treated Clay Fibro-ciment renforcé argile traitée Xiao H.W., Lee F.H., Zhang M.H., Yeoh S.Y.

2633

Large-scale Piled Raft with Grid-Form Deep Mixing Walls on Soft Ground Comportement en vraie grandeur d’une fondation mixte radier-pieux établie dans un sol meuble amélioré par quadrillage de mélange profond de sol Yamashita K., Wakai S., Hamada J.

2637

Initial investigation into the carbonation of MgO for soil stabilisation Premières investigations sur la carbonatation de MgO utilisé pour la stabilisation des sols Yi Y.L., Liska M., Unluer C., Al-Tabbaa A.

2641

Innovative solutions in the field of geotechnical construction and coastal geotechnical engineering under difficult engineering-geological conditions of Ukraine Solutions innovantes dans le domaine de la construction géotechnique et de la géotechnique côtière dans des conditions géotechniques complexes en Ukraine Zotsenko M., Vynnykov Y., Doubrovsky M., Oganesyan V., Shokarev V., Syedin V., Shapoval, Poizner M., Krysan V., Meshcheryakov G.

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Technical committee 212 Deep Foundations Comité technique 212 Fondations profondes General Report for the Two Sessions of TC 212 - Deep Foundations Rapport général des deux sessions du TC 212 - Fondations profondes Katzenbach R., Choudhury D., Chang D.W.

2651

Structural and geotechnical design of a piled raft for a tall building founded on granular soil Conception géotechnique et structurelle du radier sur pieux d’un bâtiment de grande hauteur fondé sur des sols granulaires Allievi L., Ferrero S., Mussi A., Persio R., Petrella F.

2659

Performance of Piled-Raft System under Axial Load Performance du système radier pieux sous chargement axial Alnuiam A., El Naggar H., El Naggar M.H.

2663

Analysis of Full-Scale Random Vibration Pile Tests in Soft and Improved Clays Analyses à grande échelle de vibrations aléatoires sur pieux dans un sol argileux Ashlock J.C., Fotouhi M.K.

2667

A Design Method For Piled Raft Foundations Méthode de conception des fondations de type radier sur pieux Balakumar V., Oh E., Bolton M., Balasubramaniam A.S.

2671

A practical method for the non-linear analysis of piled rafts Une méthode d’analyse pratique pour déterminer la réponse non linéaire des fondations mixtes de type radier sur pieux Basile F.

2675

A Variational Approach for Analysis of Piles Subjected to Torsion Une approche variationnelle pour l’analyse des pieux soumis à torsion Basu D., Misra A., Chakraborty T.

2679

Ancrage des pieux tarière creuse type III dans des terrains indurés : nécessité d’outils de forage performants et de reconnaissances de sols adaptées Anchoring of continue flight auger piles in hard soil : necessity of succesful tools of drilling and adapted soils investigations. Berthelot P., Cardona G., Glandy M., Durand F.

2683

Improved Safety Assessment of Pile Foundations Using Field Control Methods Évaluation améliorée de la sécurité des fondations sur pieux à l’aide de méthodes de contrôle in situ Bilfinger W., Santos M.S., Hachich W.

2687

Three Dimensional Finite Element Nonlinear Dynamic Analysis of Full-Scale Piles under Vertical Excitations Analyse dynamique non linéaire en 3D par éléments finis des pieux à grande échelle soumis à des vibrations verticales Biswas S., Manna B.

2691

P-Y curves from the prebored pressuremeter test for laterally loaded single piles Courbes P-Y à partir de l’essai pressiométrique préforé pour les pieux isolés sous charge latérale Bouafia A.

2695

Comparaison des règlements australien et français pour le dimensionnement des pieux - Prise en compte des essais de chargement French and Australian Pile Design Comparison – Load Testing Influence on Design Bretelle S.

2699

Dynamic Pile Testing at the Mesa A Rail Bridge Analyse dynamique d’essais de pieux au pont ferroviaire Mesa A Cannon J.G.

2703

Uplift behavior of bored piles in tropical unsaturated sandy soil Comportement en traction de pieux forés en sol tropical sablonneux non saturé Carvalho de D., Rocha de Albuquerque P.J.

2707

Essais de chargement statique de pieux en bois instrumentés avec des extensomètres amovibles Timber pile load test instrumented with removable extensometers Christin J., El Kouby A., Reiffsteck P., Rocher-Lacoste F.

2711

Pylon foundation of a cable stayed bridge at the motorway ring road of Wrocław Fondation d’un pylône du pont suspendu du périphérique de l’autoroute de Wrocław Dembicki E., Cudny M., Krasiński A., Załęski K.

2715

Consolidating Soil-Pile Interaction Interaction pieux-sol en cours de consolidation El-Sakhawy N., Nassar A.

2719

XXXVIII

Contents / Table des matières

The Performance of Helical Pile Groups Under Compressive Loads: A Numerical Investigation Performance d’un groupe de piles héliocoïdales sous chargement axial : une étude numérique Elsherbiny Z., El Naggar M.H.

2723

Contributing factors on soil setup and the effects on pile design parameters Facteurs contribuant au durcissement du sol et leur effet sur les paramètres de conception des pieux Fakharian K., Attar I.H., Sarrafzadeh A., Haddad H.

2727

Model loading tests in large soil tank on group behavior of piles Essais de chargement modèle afin d’étudier le comportement de groupe de pieux dans un grand réservoir du sol Goto S., Aoyama S., Liu B., Towhata I., Takita A., Renzo A.A.

2731

Probabilist analysis of the foundation of a shopping center in Brazil Analyse probabiliste des fondations d’un centre commercial au Brésil Gusmão A., Oliveira P., Ferreira S., Maia G., Amorim M.

2735

Bearing capacity of displacement piles in layered soils with highly diverse strength parameters Capacité portante des pieux de deplacements battus dans les sols stratifiés avec des paramètres fortement differés de la resistance Gwizdala K., Krasinski A.

2739

Practical experience with piled raft design for tall buildings Expérience pratique de la conception de radiers sur pieux pour les immeubles de grandes hauteurs Haberfield C.M.

2743

Non-Conventional Pile Loading Tests in Vietnam Essai non conventionnel de chargement de pieux au Vietnam Hai N.M., Dao D.H.

2747

Slope stability structures for road landslide Structures de stabilité de pentes pour glissement de terrain Hamova M., Frangov G., Zayakova Hr.

2751

Research on the Load-Bearing Behaviour of Bored Piles with Different Enlarged Bases La recherche sur le comportement portante de pieux forés avec diverses bases élargies Herrmann R.A., Löwen M., Tinteler T., Krumm S.

2755

Visualization of Settlement Behavior for Friction Pile Group during Consolidation Visualisation du tassement pour un groupe de pieux frottant lors d’une consolidation Ishikura R., Matsuda H., Igawa N.

2759

Interactive 3-D Analysis Method of Piled Raft Foundation for High-rise Buildings Méthode d’analyse 3-D interactive de fondations mixte radier pieux pour immeubles de grande hauteur Jeong S.J, Cho Ja.

2763

Optimal FBG Sensor Deployment via Gaussian Quadrature Formula for Measurement of Displacement of Laterally Loaded Piles Le déploiement optimal des capteurs à fibres optiques, par la formule de la quadrature de Gauss, pour la mesure du déplacement des pieux chargés latéralement Jung Y.-H., Na S.-U., Mok Y. Numerical Simulation of the Load Tests on Bearing Capacity of Piled Raft Foundations Simulations numériques d’essais de chargement pour établir la capacité portante des fondations mixtes radier sur pieux Kaneda K., Honda T., Shigeno Y., Hamada J. The Development and the Structural Behavior of a New Type Hybrid Concrete Filled Fiber-Glass Reinforced Plastic Pile Développement et comportment structural d’un nouveau type de béton hybride rempli de fibre de verre renforcé par pile plastique Kang I.-K., Kim H.-T., Baek S.-C., Park S.-Y.

2767

2771

2775

Ground displacements related to deep excavation in Amsterdam Déformations du sol liées à des excavations profondes à Amsterdam Korff M., Mair R.J.

2779

Drilled pile technology in retaining wall construction and energy transfer Application de la technologie des pieux forés à la construction des murs de soutènement et au tranfert d´énergie Lehtonen J.

2783

Three-Dimensional Models of Bearing Capacity - Case Study Modèles tridimensionnels de capacité de portante - Étude de cas Leite da Silva C.P., Moreira de Souza N., Medeiros Silva C.

2787

Full scale rapid uplift tests on transmission tower footings Tests grandeur nature d’arrachement rapide sur les fondations d’une tour relais Levy F.M., Richards D.J.

2791

XXXIX

Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Characteristics Values in Rock Socket Design Valeurs caractéristiques d’ancrage sur roche Look B., Lacey D.

2795

Safety theory in geotechnical design of piled raft Théorie sur la sécurité pour la réalisation de radier sur pieux Lorenzo R., Zubeldia E.H., Cunha R.P.

2799

Bored pile foundation response using seismic cone test data Réponse des pieux à l’aide des données de piézocône sismique Mayne P.W., Woeller D.J.

2803

Energy and Reliability Applied to Continuous Flight Augern Pilings - The SCCAP Methodology Énergie et fiabilité appliquées à l’excavation des pieux forés en continu - La méthodologie SCCAP Medeiros Silva C., Camapum de Carvalho J., Brasil Cavalcante A.L.

2807

Performance of a pioneer foundation of the skirt type for the Metro-Line 12 overpass on the Mexico City soft clay Comportement d’un nouveau type de fondations de type radier à jupe, utilisé pour les tronçons en viaduc de la ligne 12 du Métro fondés sur les argiles molles de Mexico Mendoza M.J., Rufiar M., Ibarra E., Mendoza S.A.

2811

Improving the capacity of bored piles by shaft grouting Améliorer la capacité portante des pieux forés par injection de coulis opérée latéralement Miller M., Potts V., Skinner H., Vaziri M.

2815

Polymer pillar, a new innovation for underpinning Colonne de polymère, une nouvelle innovation comme support de fondation Perälä A.

2819

Identification of Test Pile Defects in a Super-tall Building Foundation Identification des anomalies dans les essais de chargement de pieu pour les fondations d’une tour de très grande hauteur Poulos H.G., Badelow F., Tosen R., Abdelrazaq, Kim S.H.

2823

A review of pile test results and design from a London clay site Un compte rendu sur les resultants d’essais sur pieux et leur dimensionnement sur un site d’argile de Londres Powell J.J.M., Skinner H.

2827

Effet du mode de mise en place sur le comportement statique de pieux dans l’argile fortement surconsolidée des Flandres Effect of installation mode on the static behaviour of piles in highly overconsolidated Flanders clay Puech A., Benzaria O.

2831

Analysis of Piles Supporting Excavation Adjacent to Existing Buildings Analyse de pieux de bâtiments existant en cours de fouilles sous-jacentes Ramadan E.H., Ramadan M., Khashila M.M., Kenawi M.A.

2835

Analysis and Design of Piles for Dynamic Loading Analyse et conception de fondations par pieux en chargement dynamique Ray R.P., Wolf Á.

2839

A new tool for the automated travel time analyses of bender element tests Un nouvel outil pour les analyses automatisées du temps de déplacement des essais « bender element » Rees S., Le Compte A., Snelling K., Rinaldi V.A., Viguera R.

2843

Pseudo-static Pile Load Test: Experience on Pre-bored and Large Diameter Piles Tests de chargement pseudo-statique sur pieux: experiences sur pieux forés de grands diamètres Rinaldi V.A., Viguera R.

2847

Behavior of Vertical Piles Embedded in Sand under Inclined Loads near Ground Slope Comportement de pieux verticaux ancrés dans une couche de sable à proximité d’une pente Sakr M.A., Nasr A.M.

2851

Semi-Analytical Solutions for Laterally Loaded Piles in Multilayered Soils Solutions Semi-analytiques pour des pieux soumis à des charges latérales dans les sols multicouches Salgado R, Basu D., Prezzi M, Tehran F.S.

2855

Skyscrapers of «Moskva-City» Business Center - Tests of Bored Piles Gratte-ciel du centre d’affaires « Moskva-City » – Essais de pieux forés Shulyatiev О.А., Ladyzhensky I.G. Yastrebov P.I.

2859

Cavity remediation for pylon foundation of the Transrhumel Viaduct in Constantine Résolution des problèmes de cavité sous les fondations du Viaduc Trans-Rhumel de Constantine Steenfelt J.S., Schunk M.

2863

Integrating Nonlinear Pile Behavior with Standard Structural Engineering Software Analyse non linéaire de fondations par pieux à l’aide d’un code industriel Szép J., Ray R.P.

2869

XL

Contents / Table des matières

Experimental Study on the Method of Rebound and Recompression Deformation Calculation in Deep and Large Foundation Design Etude expérimentale sur la méthode de calcul des déformations de résilience et de recompression pour les fondations larges et profondes Teng Y., Li J., Wang S. Deep Basement Construction of Bank of Thailand Along Chao Phraya River closed to Tewavej Palace and Bangkhumphrom Palace Construction du sous-sol profond de la Banque de Thaïlande le long de la Chao Phraya près des palais de Bangkhumphrom et Tewavej Teparaksa W.

2873

2877

Creep and long-term bearing capacity of a long pile in clay Fluage et capacité portante à long terme d’un long pieu dans de l’argile Ter-Martirosyan Z.G., Ter-Martirosyan A.Z., Sidorov V.V.

2881

Compressive resistance of piles, an update Résistance à la compression des pieux, une mise à jour Tol van A.F., Stoevelaar R., Bezuijen A., Jansen H.L., Hannink G.

2885

A design verification method for pile foundations used in combination with solidified improved columns Une méthode de vérification de la conception des pieux en combinant avec des colonnes de sol améliorés Tomisawa K., Miura S.

2889

nfluence of multiple helix configuration on the uplift capacity of helical anchors Influence de la configuration des hélices sur la résistance à l’arrachement de pieux hélicoïdaux Tsuha C.H.C., Santos T.C., Rault G., Thorel L., Garnier J.

2893

Super-long bored pile foundation for super high-rise buildings in China Fondation profonde sur pieux de très grandes longueurs pour les immeubles de grandes hauteurs en Chine Wang W., Wu J., Li Y.

2897

Case Studies of Cost-effective Foundation Design in Rock Études de cas sur la conception de la Fondation rentable dans Rock Wong P.K.

2901

Difficulté d’exécution des pieux profonds de grand diamètre dans des sols mous Difficulty execution of large diameter deep piles in soft soils Zaghouani K., Chouikha A., Haffoudhi S.

2905

Load Tests on Full-Scale Bored Pile Groups Essais de chargement sur des groupes de pieux forés Zhang Y., Salgado R., Dai G., Gong W.

2909

Technical committee 214 Foundation Engineering for Difficult Soft Soil Conditions Comité tevnique 214 Fondations en conditions difficiles de sols mous General Report of TC 214 - Soft soils Rapport général du TC 214 - Sol mous Ovando-Shelley E., Rangel-Núñez J.L.

2915

Soil Fracturing Induced by Land Subsidence in Mexico City Fracturation des sols induite par la subsidence de la ville de Mexico Auvinet G., Méndez E., Juárez M.

2921

Characterization of Sensitive Soft Soils for the Waterview Connection Project, New Zealand Caractérisation de sols mous sensibles pour le projet de raccordement Waterview en Nouvelle-Zélande Bobei D.C., Locks J.

2925

Design and Construction of a Landfill Containment Bund cum Seawall Supported on Stone Columns Installed in Very Soft Marine Mud in Cotai, Macau Conception et construction d’un remblai de depôts avec une enceinte sur des colonnes ballastées nstallées dans un sol marin très mou à Cotai, Macao De Silva S., Fong L.T.T.

2929

Estimation of undrained shear strength of soft soil obtained by cylinder vertical penetration Estimation de la résistance au cisaillement d’un sol mou en conditions non-drainées obtenue par la pénétration verticale d’un cylindre Equihua-Anguiano L.N., Orozco-Calderon M., Foray P.

2933

he Application of a Novel Design Approach for Construction over soft soils: The Hybrid Undrained-Drained model L’application d’une nouvelle méthode de conception pour des constructions sur sols mous: le modèle hybride non drainé - drainés Espinoza D., Li C.

2937

XLI

Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Land reclamation on soft clays at Port of Brisbane Construction d’un terre-plein sur des sols argileux dans le port de Brisbane Ganesalingam D., Sivakugan N., Ameratunga J., Schweitzer G.

2941

Kansai International Airport. Theoretical settlement history Aéroport international de Kansai. Historique théorique du tassement Juárez-Badillo E.

2945

Design and Performance of Highway Embankments Constructed Over Sri Lankan Peaty Soils Conception et performance de remblais d’autoroute construits sur sols tourbeux au Sri Lanka Karunawardena A., Toki M.

2949

Design improvements for expansion of a roadway on a thick layer of soft soil Un projet d’amélioration pour l’élargissement d’une autoroute sur une argile molle Kim K.-H., Jung T.-M., Jung J.-H., Kim T.-H., Kim S.-R., You S.-H.

2953

Case Study on X-section Cast-in-place Pile-Supported Embankment over Soft Clay Étude de cas pour un remblai renforcé par des pieux de section en X coulés en place dans de l’argile molle Liu H.L., Kong G.Q., Ding X.M, Yu T., Yang G.

2957

Settlements of Earth Fills on Thick Layers of Overconsolidated Soft Clays without Geodrains Tassements des remblais sur d’épaisses couches d’argile molle, surconsolidée, sans géodrains Massad F., Teixeira A.H., Carvalho C.T., Grangé L.F.A.

2961

Aspects on the modelling of smear zones around vertical drains Aspects de la modélisation de la zone remaniée autour des drains verticaux Müller R., Larsson S.

2965

A Review of Geogrid Working Platform in Soft Ground in Malaysia Analyse du comportement de plateformes renforcées par géogilles en Malaisie Ooi T.A., Tee C.H., Chan C.B., Ong R.

2969

Container Terminal on Soft Soil Terminal de conteneurs sur un sol mou Popovic N., Stanic B.

2973

Instrumented Trial Embankment on Soft Ground at Tokai, State of Kedah, Malaysia Embankment essai instrumenté sur un sol mou, État de Kedah, Malaisie Tan Y.C., Lee P.-T., Koo K.-S

2977

Prediction of and countermeasures for embankment-related settlement in ultra-soft ground containing peat Prédiction et contre-mesure sur les tassements de remblais dans les sols ultra-meubles contenant de la tourbe Tashiro M., Inagaki M., Asaoka A.

2981

Numerical simulation of energy consumption of artificial ground freezing applications subject to water seepage Simulation numérique de la consummation d’énergie des applications pour la congélation artificielle du sol soumise au flux de l’eau souterraine Ziegler M., Schüller R., Mottaghy D.

2985

Technical committee 215 Environmental Geotechnics Comité technique 215 Géotechnique de l’environnement General Report of TC 215 - Environmental Geotechnics Rapport Général du TC 215 - Géotechnique de l’Environnement Bouazza A.

2991

Novel bentonites for containment barrier applications Bentonites novatrices pour des applications comme barriers de confinement Bohnhoff G., Shackelford C., Malusis M., Scalia J., Benson C., Edil T., Di Emidio G., Katsumi T., Mazzieri F.

2997

Long term performance of cement-bentonite cut-offs in saline and acidic solutions Perméabilité à long terme des parois ciment-bentonite en solutions acides et salines Brianzoni V., Fratalocchi E., Pasqualini E.

3001

Determination of shear strength of MSW. Field tests vs. laboratory tests Détermination de la résistance au cisaillement des déchets urbains (MSW). Essais in situ vs essais de laboratoire Cañizal J., Lapeña P., Castro J., Costa da A., Sagaseta C.

3005

Geo-environmental problems in landfills of MSW with high organic content Problèmes géo-environnementaux dans les sites d’enfouissement de déchets urbains à hautes teneurs organiques Chen Y.M., Zhan L.T., Xu X.B., Liu H.L.

3009

Étude expérimentale d’une technique de filtration radiale pour une application au sein de Barrières Perméables Réactives (BPR) Experimental study of radial filtration in Permeable Reactive Barriers (PRB) Courcelles B.

XLII

3013

Contents / Table des matières

Measurement of NAPL saturation distribution in whole domainsby the Simplified Image Analysis Metod Mesure de la distribution de la satturation de liquide en phase non aqueuse couvrant tout le spectre de l’étude par la méthode simplifiée d’analyse d’image Florès G., Katsumi T., Inui T., Takai A.

3017

Hydraulic conductivity of zeolite-sand mixtures permeated with landfill leachate Conductivité hydaulique de mélanges zéolithe-sable infiltrés par des écoulements de décharge de déchets Fronczyk J., Garbulewski K.

3021

Moisture-Suction Relationships for Geosynthetic Clay Liners Courbes de rétention des membranes géotextiles chargées en argile Hanson J.L., Risken J.L., Yeşiller N.

3025

Hydraulic conductivity of compacted clay liners moisture-conditioned and permeated with saline coal seam gas water La conductivité hydraulique de l’humidité argile compactée doublures conditionné et imprégné avec de l’eau salée gaz de houille couture Indrawan I.G.B., Williams D.J., Scheuermann A. Simultaneous estimation of transverse and longitudinal dispersion in unsaturated soils using spatial moments and image processing Estimation simultanée de la dispersion transversale et longitudinale dans des sols insaturés au moyen de la méthode des moments pour l’analyse des données spatiales et du traitement d’images Inoue K., Shimada H., Tanaka T.

3029

3033

Evaluating the long-term leaching characteristics of heavy metals in excavated rocks Évaluation des caractéristiques de lixiviation à long terme de métaux lourds dans les roches excavées Inui T., Katsumi T., Takai A., Kamon M.

3037

Geo-environmental challenges of a major coal terminal development in Australia Défis géo-environnementaux du développement d’un terminal majeur de charbon en Australie Jones S.R.

3041

Characterisation of landfill steel mill sludge waste in terms of shear strength, pore water pressure dissipation and liquefaction potential Caractérisation de la résistance au cisaillement, de l’évolution des pressions d’eau interstitielle et du potentiel de liquéfaction des boues d’aciérie dans un centre de stockage. Lavoie J.L.N., Sinclair T.J.E.

3045

A numerical analysis of phytoextraction processes Une analyse numérique des processus de phyto-extraction Lugli F., Mahler C.F.

3049

Soil-geosynthetic interface strength on smooth and texturized geomembranes under different test conditions Résistance au cisaillement des interfaces entre sols et membranes géo-synthétiques lisses ou rugueuses sous différentes conditions Monteiro C.B., Araújo G.L.S., Palmeira E.M., Cordão Neto M.P.

3053

Geoenvironmental Approach to Restoration of Agricultural Land Damaged by Tsunami Approche géo-environnementale de la restauration de terres agricoles endommagées par Tsunami Omine K., Moqsud M.A., Hazarika H.

3057

Factors affecting hydration of Geosynthetic Clay Liners in landfill applications Facteurs influençant l’hydratation des géosynthétiques bentonitiques dans les applications d’enfouissement Rayhani M.T., Sarabadani H.

3061

Utilisation de la désorption thermique pour l’élimination in situ des couches flottantes d’hydrocarbures Use of thermal desorption for removing in-situ floating oil layers Saadaoui H., Haemers J., Denecheau P., Cédou C.

3065

Devepment and Verification of Ecohabitat Chart based on Ecological Geotechnics Développement et vérification du diagramme Ecohabitat, basé sur la géotechnique écologique Sassa S., Watabe Y., Yang S.

3069

A New Approach for Characterizing Shear Strength of Municipal Solid Waste for Land Fill Design Une nouvelle approche pour la caractérisation de la résistance au cisaillement des déchets urbains pour la conception des décharges Singh S.

3073

The role of molecular biology in geotechnical engineering Le rôle de la biologie moléculaire en géotechnique Stewart D.I., Fuller S.J., Burke I.T., Whittleston R.A., Lockwood C.L., Baker A.

3077

A System of dehydration, purification, and reduction for dredged soil – Release inhibition of nutrient salts from bed mud using natural zeolite Un système de déshydratation, d’épuration et de réduction de sols dragués - Prévention du relâchement de sels nutritifs des lits de boue à l’aide d’une zéolithe naturelle Umezaki T., Kawamura T.

XLIII

3081

Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Technical committee 301 Preservation of Historic Sites Comité technique 301 Préservation des sites historiques General Report du TC 301 - Monuments, historic sites and case histories Rapport général du TC 301 - Monuments, sites historiques et études de cas Flora A. Reconstitution of foundation platform of Prasat Suor Prat by compaction of original soil with slaked lime, Angkor Ruins, Cambodia Reconstitution de la plate-forme de la fondation de Prasat Suor Prat par compactage du sol d’origine additionné de chaux éteinte, sur les ruines d’Angkor, au Cambodge Akazawa Y., Fukuda M., Iwasaki Y., Nakazawa J.

3087

3095

Geotechnical Aspects of Design and Construction of the Mountain Cluster Olympic Facilities in Sochi Les aspects géotechniques des projets et de la construction des sites olympiques situés dans les pays montagneux autour de la ville Sotchi Fedorovsky V., Kurillo S., Kryuchkov S., Bobyr G., Djantimirov K., Iliyn S., Iovlev I., Kharlamov P., Rytov S., Skorokhodov A., Kabantsev О.

3099

Importance of understanding the development and significance of sulphates in the London Clay L’importance de comprendre le développement et la signification des sulfates dans l’Argile de Londres. Hawkins A.B., St John T.W.

3103

Rockfall-protection embankments – design concept and construction details Merlons de protection contre les chutes de pierres - modèle de conception et d’exécution Hofmann R., Vollmert L., Mölk M.

3107

Authenticity of Foundations for Heritage Structures Authenticité des fondations pour les structures du patrimoine Iwasaki Y., Zhussupbekov A., Issina A.

3111

Geotechnical Assessment for the Restoration of Garandoya tumulus with the Naked Stone Chamber Évaluation géotechnique de la restauration du tumulus de Garandoya et grottes en pierres nues Mimura M., Yoshimura M.

3115

Geotechnical Features of Sochi Olympic Facilities Project Designs Les aspects géotechniques de la conception des installations olympiques de Sochi Petrukhin V.P., Kolybin I.V., Budanov V.G., Isaev О.N., Kisin B.F., Bokov I.A.

3119

Heaving Mechanisms in High Sulfate Soils Mécanismes de soulèvement dans les sols à contenu élevé en sulfates Puppala A.J., Talluri N., Gaily A., Bhaskar, Chittoori C.S.

3125

Geotechnical aspects in sustainable protection of cultural and historical monuments Les aspects géotechniques de la protection durable des monuments culturels et historiques Sesov V., Cvetanovska J., Edip K.

3129

Modern methods of geotechnical defense of buildings in the difficult geological conditions of Ukraine Méthodes modernes pour la défense géotechnique de bâtiments dans les conditions géologiques difficiles de l’Ukraine Slyusarenko Y., Chervinskyy Y., Karpenko Y., Dvornik S., Malikov S., Rozenvasser G., Lavshuk I.

3133

Geotechnical problems related to the development of territories in the conditions of the Republic of Tajikistan Problèmes géotechniques lies au développement de territoires dans les conditions de la République du Tadjikistan Usmanov R.A., Saidov R.S., Mangushev R.A.

3137

The preservation of Agrigento Cathedral La conservation de la cathédrale d’Agrigente Valore C., Ziccarelli M.

3141

Geotechnical characteristics of glacial soil deposits at Punta Arenas in Chilean Patagonia Caractéristiques géotechniques des dépôts glaciaires du sol à Punta Arenas en Patagonie chilienne Vásquez A., Le Roux J.-P., Foncea C.

3145

Geotechnical Issues of Megaprojects on Problematical Soil Ground of Kazakhstan Questions géotechniques de mégaprojets sur sol problématique du Kazakhstan Zhussupbekov A.Zh., Ling H.I., Baitassov T.M., Lukpanov R.E., Tulebekova A.S., Yenkebayev S.B., Popov V.N., Krasnikov S.V., Boominathan A.

3149

XLIV

Contents / Table des matières

Technical committee 307 Sustainability in Geotechnical Engineering Comité technique 307 Construction durable en géotechnique General Report of TC 307 - Sustainability in Geotechnical Engineering Rapport général du TC 307 - Durabilité en géotechnique Basu D., Puppala A.J., Chittoori B.

3155

Evaluation of Rubber/Sand Mixtures as Replacement Soils to Mitigate Earthquake Induced Ground Motions Évaluation du mélange sable-caoutchouc comme sol de remplacement pour atténuer les mouvements sismiques Abdelhaleem A.M., El-Sherbiny R.M., Lotfy H., Al-Ashaal A.A.

3163

New Replacement Formations on Expansive Soils Using Recycled EPS Beads Remplacement sur les sols expansifs en utilisant des perles EPS Abdelrahman G.E., Mohamed H.K., Ahmed H.M.

3167

Sustainability in Geotechnical Engineering Viabilité en géotechnique Basu D., Misra A., Puppala A.J., Chittoori C.S.

3171

Mechanics of Manufactured Soil Using Powder Wastes Mécanique des sols fabriqués à partir de déchets de poudre Baykal G.

3175

Méthodes non traditionnelles de traitement des sols : apports techniques et impact sur le bilan environnemental d’un ouvrage en terre Soil treatment with non traditional additives in earthworks: evaluation of the technical and environmental improvements Blanck G., Cuisinier O., Masrouri F.

3179

Advanced testing and modelling delivers cost effective piled raft foundation solution Essais avancés et modélisation délivre une solution économique empilés fondation sur radier Bourne-Webb P., Cunningham M., Card G.

3183

The use of Recycled Aggregates in Unboud Road Pavements L’utilisation d’aggégats recyclés en revêtements de chaussée sans liant Cameron D.A., Rahman M.M., Azam A.M., Gabr A.g., Andrews R., Mitchell P.W.

3187

Reuse of dredged sediments for hydraulic barrires : adsorption and hydraulic conductivity improvement through polymers La réutilisation des sédiments dragués pour barrières htdrauliques : l’adsorption et l’améloration de la conductivité hydraulique avec des polymères Di Emidio G., Verastegui Flores R.D., Bezuijen A.

3191

Characterization of recycled materials for sustainable construction Caractérisation des matériaux pour la construction drable Edil T.B.

3195

Technical and Economic Analysis of Construction and Demolition Waste Used in Paving Project Analyse technique et économique des déchets dans la construction de pavage Farias A., Fucale S., Gusmão A., Maia G.

3199

Comparative Life Cycle Assessment of Geosynthetics versus Conventional filter layer Analyse de cycle de vie comparative d’une couche de filtre géotextile et conventionnelle Frischknecht R., Büsser-Knöpfel S., Itten R., Stucki M., Wallbaum H.

3203

La réutilisation des fondations existantes dans les projets de réhabilitation de constructions anciennes Reuse of existing foundations for the rehabilitation of old buildings Guilloux A., Le Bissonnais H., Saussac L., Perini T.

3207

Modern geotechnical construction methods for important infrastructure buildings Méthodes de construction modernes des ouvrages géotechniques dans les grands projects d’infrastrcuctures Heerten G., Vollmert L., Herold A., Thompson, Dupond J., Alcazar G.

3211

Sustainable Management of Contamined Sediments Gestion durable des sédiments contaminés Holm G., Lundberg K., Svedberg B.

3215

Polymer support fluids: use and misuse of innovative fluids in geotechnical works Les polymères: l’utilisation de nouveaux fluides de forage en travaux géotechnique Jefferis S.A., Lam C.

3219

Utilisation of polyethylene (plastic) shopping bags waste for soil improvement in sandy soils Utilisation des déchets de sacs en polyéthylène (plastiques) pour l’amélioration des sols sableux Kalumba D., Chebet F.C.

3223

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Effect of dredge soil on the strength development of air-foam treated lightweight soil Effets des sols de dragage sur le développement de la résistance des sols mélangés à de l’air Kataoka S., Horita T., Tanaka M., Tomita R., Nakajima M.

3227

Application of a Method to Accelerate Granulated Blast Furnace Slag Solidification Une méthode de solidification accélérée des granulats issus de laitier de haut fourneau Kikuchi Y., Mizutani T.

3231

Building on an old landfill: design and construction Construire sur une ancienne décharge : dimensionnement et exécution des travaux McIntosh G.W., Barthelmess A.J.

3235

Interpretation of mechanical behavior of cement-treated dredged soil based on soil skeleton structure Interprétation des comportements mécaniques des sols dragués traités au ciment basée sur la structure squelette du sol Nakano M., Sakai T.

3239

Utilization of waste copper slag as a substitute for sand in vertical sand drains and sand piles Utilisation des scories de cuivre en tant que substitut pour le sable dans le sable drains verticaux et tas de sable Nawagamuwa U.P., Senanayake A., Rathnaweera T.

3243

Tools for Natural Hazard management in a Changing Climate Outils de gestion de désastres naturels dans un climat changeant Rogbeck Y., Löfroth H., Rydell B., Andersson-Sköld Y.

3247

Experimental reinforced soil walls built with recycled construction and demolition waste (RCDW) Murs expérimentaux de sol renforcé construits avec résidus de construction et démolition recyclés Santos E.C.G., Palmeira E.M.

3251

Comparing the properties of EPS and glass foam mixed with cement and sand Comparer les propriétés d’EPS et mousse de verre mélangé avec du ciment et du sable Teymur B., Tuncel E.Y., Ahmedov R.

3255

Geotechnical engineering and protection of environment and sustainable development Engineering géotechnique, protection de l’environnement et développement durable Vaníček M., Jirásko D., Vaníček I.

3259

Applicability of Municipal Solid Waste (MSW) Incineration Ash in Road Pavements Base Utilisation de cendres d’incinération de déchets solides municipaux (MSW) dans la couche de base de chaussée Vizcarra G., Szeliga L., Casagrande M., Motta L.

3263

Research Results of Fine-Grained Soil Stabilization Using Fly Ash from Serbian Electric Power Plants Les résultats de recherche de la stabilisation des sols de grains fins en utilisant les cendres volantes des centrales électriques serbes Vukićević M., Maraš-Dragojević S., Jocković S., Marjanović M., Pujević V.

3267

Simplified Prediction of Changes in Shear Strength in Geotechnical Use of Drinking Water Sludge Prédiction simplifiée de changements dans la force du ciseau dans usage Geotechnical de boue de l’eau potable Watanabe Y., Komine H.

3271

Road foundation construction using lightweight tyre bales Construction des assises de routes à l’aide de balles de pneus légères Winter M.G.

3275

Technical committee 210 + 201 Dykes Leeves and Dams Comité technique 210 + 201 Digues, levées et barrages General report - Geotechnical problems of dikes (TC 201) and dams (TC 210) Rapport général - Problèmes géotechniques dans les digues (TC 201) et barrages (TC 210) Xu Z.

3281

Hydraulic failure of flood protection dykes Défaillance du circuit hydraulique des levées de protection contre les inondations Brandl H., Szabo M.

3289

Prédiction du comportement de barrage en enrochement de grande taille à l’aide d’une modélisation tridimensionnelle 3293 Prediction of the behavior of very high CFRD using a 3D modelling Chen Y., Fry J.-J., Laigle F., Vincens E., Froiio F. Slope stability of the Włocławek Dam frontal earth dam in the light of the modernisation works carried out in the period 2000-2011 Stabilité de la pente du barrage en terre de Włocławek à la lumière des travaux de modernisation exécutés dans la période 2000-2011 Leszczynski M., Lipiecki B., Popielski P.

XLVI

3297

Contents / Table des matières

Deformation safety of high concrete face rockfill dams Calculs en déformations de la sécurité des grands barrages en rochement à masque amont en béton Li N., Wang J., Mi Z., Li D. Safety of a protection levee under rapid drawdown conditions. Coupled analysis of transient seepage and stability La sécurité d’une digue de protection en conditions de vidange rapide. Analyse couplée des écoulements transitoires et de la stabilité López-Acosta N.P., Fuente de la H.A., Auvinet G.

3301

3305

Some Technical Aspects of the Tailing Dam Failure at the Ajka Red Mud Reservoirs Quelques aspects techniques de la rupture d’une digue de retenue de boues à Ajka Mecsi J.

3309

The Design of Filter Materials and their Importance in Geotechnical Engineering La conception de matériaux filtrants et leur importance en géotechnique Messerklinger S.

3313

Identification du risque d’érosion interne sur les digues de l’Isère et du Drac Identification of erosion risk on the Isère and Drac river levees Monnet J., Plé O., Nguyen D.M.

3317

Suffusion in compacted loessial silts. Interaction with granular filters Suffusion dans les limons lœssique compactés. Interaction avec les filtres granulaires Terzariol R.E., Rocca R.J., Zeballos M.E.

3321

Predicting long-term settlements of coastal defences for the safeguard of the Venetian Lagoon Évaluation des tassements de consolidation secondaire des structures côtières de protection pour la sauvegarde de la lagune de Venise Tonni L., García Martínez M.F., Simonini P.

3325

Full scale field tests for strength assessment of peat Essais in situ en vraie grandeur pour évaluer la résistance d’une tourbe Zwanenburg C., Van M.A.

3329

Technical committee 307 + 212 Heat effects Comité technique 307+212 Effets de la cHAleur General Report TCs 307+212 - Thermal Geomechanics with Emphasis on Geothermal Energy Rapport général TCs 307+212 - Géomécanique thermique avec une attention particulière portée sur l’énergie géothermique Puppala A.J., Choudhury D., Basu D. Numerical Modelling of Ground Heat Exchangers with Different Ground Loop Configurations for Direct Geothermal Applications Modélisation numérique des échangeurs de chaleur souterrains avec différentes configurations de boucles pour les applications géothermiques directs Bidarmaghz A., Narsilio G., Johnston I.

3335

3343

The response of energy foundations under thermo-mechanical loading La réponse des fondations thermo actifs sous chargement thermo-mécanique Bodas Freitas T.M., Cruz Silva F., Bourne-Webb P.J.

3347

Large Thermal Energy Storage at Marstal District Heating Importante capacité de stockage de l’énergie thermique pour le chauffage collectif de Marstal Dannemand Andersen J., Bødker L., Jensen M.V.

3351

Combination of borehole heat exchangers and air sparging to increase geothermal efficiency Combinaison de sondes géothermiques et barbotage d’air pour augmenter l’efficacité géothermique Grabe J., Menzel F., Ma X.

3355

Geothermal Heat PipeBorehole Heat-Exchangers: Computational Simulation and Analysis of Measurement Data Échangeurs thermiques à thermosiphon utilisés en géothermie : simulation numérique et analyse des mesures Katzenbach R., Clauss F.

3359

Analysis of the freeze thaw performance of geothermal heat exchanger borehole grout materials Étude de la résistance au gel et dégel des sondes géothermiques verticales Katzenbach R., Clauss F., Rochée S.

3363

Thermal influences on swelling pressure and swelling deformation of bentonites and investigation of its factors Effets thermiques sur la pression et les déformations de gonflement des bentonites et facteurs d’influence Komine H.

3367

Performance of Piled Foundations Used as Heat Exchangers Performance des fondations sur pieux utilisées comme échangeurs thermiques Loveridge F., Powrie W.

3371

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Measuring soil thermal properties for use in energy foundation design La mesure des caractéristiques thermiques du sol pour la conception des fondations énergie Low J.E., Loveridge F.A., Powrie W.

3375

Thermo-Mechanical Behavior of Energy Foundations Comportement thermo-mécanique des pieux énergétiques McCartney J.S., Murphy J.S., Stewart M.A.

3379

Pressurisation thermique dans l’argile de Boom Thermal pressurization in Boom clay Monfared M., Delage P., Sulem J., Mohajerani M., Tang A.M.

3383

Effet des conditions environnementales sur les propriétés mécaniques d’un ciment de puits géothermique Effect of environmental conditions on the mechanical properties of geothermal well cement paste Nauleau E., Martineau F., Kréziak C., Ghabezloo S.

3387

Development of a predictive framework for geothermal and geotechnical responses in cold regions experiencing climate change Développement d’un cadre conceptuel pour les réponses géotechniques et géothermales dans une zone polaire sous l’influence du changement climatique Nishimura S., Jardine R.J., Fenton C.H., Olivilla S., Gens A., Martin C.J. An undrained upper bound solution for the face stability of tunnels reinforced by micropiles Une solution en limite supérieure non drainée pour la stabilité du front de tunnels renforcés par micropieux Pinyol N.M., Alonso E.E.

3391

3395

Numerical simulation of the process of geothermal low-potential ground energy extraction in Perm region (Russia) 3399 Modélisation numérique du procès de la sélection géothermale d’énergie potentielle basse du sol dans les conditions de la région de Perm (Russie) Ponomaryov A., Zakharov A. Determination of the thermal parameters of a clay from heating cell tests Détermination des paramètres thermiques d’une argile à partir d’essais dans une cellule de chauffage Romero E., Lima A., Gens A., Vaunat J., Li X.L.

3403

Analyse de la portance des pieux géothermiques Discussions about the bearing capacity of geothermal piles Suryatriyastuti M., Mroueh H., Burlon S., Habert J.

3407

One-dimensional compressive behaviour of reconstituted clays under high temperature and small strain rate Comportement oedométrique des argiles reconstituées sous fortes température et à faible vitesse de déformation Tsutsumi A., Tanaka H.

3411

Field Investgation of a geothermal energy pile: Initial Observations Essai sur site d’un pieu géothermique : observations initiales Wang B., Bouazza A., Singh R.M., Barry-Macaulay D., Haberfield C., Chapman G., Baycan S.

3415

THM simulations for laboratory heating test and real-scale field test Simulations THM d’essais de chauffage en laboratoire et en vraie grandeur in situ Xiong Y.L., Zhang F., Nishimura T., Kurimoto Y.

3419

New Developments in near-surface geothermal energy systems Nouveaux Développements dans les systèmes géothermiques proches à la surface Ziegler M., Kürten S.

3423

Understanding the effects of high temperature processes on the engineering properties of soils Comprendre les effets des procédés à haute température sur les propriétés des sols Zihms S.G., Switzer C., Karstunen M., Tarantino A.

3427

Technical committee CFMS shallow foundation Comité technique CFMS Fondations superficielles General Report - Shallow foundations Rapport général - Fondations superficielles Zerhouni M.I., Demay B.

3433

Bearing capacity of shallow foundation under eccentrically inclined load Capacité portante d’une fondation superficielle sous une charge inclinée excentrique Atalar C., Patra C.R., Das B.M., Sivakugan N.

3439

Estimating settlements of footings in sands – a probabilistic approach Estimation des tassements de semelles dans les sables – une approche probabiliste Bungenstab F.C., Bicalho K.V., Ribeiro R.C.H., Aoki R.C.H.

3443

XLVIII

Settlement velocity measured over ten years in major-scale shallow foundations on a preloaded 20-m thick silty alluvial layer Velocité des affaissements mesurés sur dix ans, sur une foundation superficielle de grandes dimensions sur une couche alluviale limoneuse de 20 m d’épaisseur préchargée Dapena E., Román F., Pardo de Santayana F., Cuéllar V.

3447

Combined massive and plate foundations under machines with dynamic loadings Des fondations combinées à blocs et plaques pour des machines avec charges dynamiques Kirichek Y., Bolshakov V.

3451

Settlements Under Footings on Rammed Aggregate Piers Tassements sous des semelles sur pieux d’agrégats battus Kuruoglu O., Horoz A., Erol O.

3455

Interaction of Nearby Strip Footings Under Inclined Loading Interaction de semelles rapprochées soumises à des charges inclinées Nainegali L.S., Ghosh P., Basudhar P.K.

3459

Over a decade of experience with computer aided learning in geotechnical engineering Plus d’une décennie d’expérience dans le domaine de l’enseignement assisté par ordinateur dans le domaine de l’ingénierie géotechnique Springman S.M., Herzog R., Seward L.

3463

Predicting Settlements of Shallow Footings on Granular Soil Using Nonlinear Dynamic Soil Properties Prédiction des tassements de fondations superficielles sur des sols granulaires en utilisant des propriétés dynamiques non linéaires du sol. Stokoe K.H., Kacar O., Van Pelt J.

3467

Characterization of Model Uncertainty in Immediate Settlement Calculations for Spread Footings on Clays Caractérisation de l’incertitude des modèles de calculs du tassement immédiat de semelles reposant sur des sols argileux Strahler A.W., Stuedlein A.W.

3471

Probalistic Assessment of the bearing Capacity of Shallow Strip Footings on Stiff-over-Soft Clay Évaluation probabiliste de la capacité portante de semelles filantes peu profondes sur couche d’argile recouvrant une couche d’argile molle Tian Y., Cassidy M.J., Uzielli M.

3475

Residual Soils and the Teaching of Soil Mechanics Les sols résiduels et l’enseignement de la mécanique des sols Wesley L.D.

3479

Application of The Tangent Modulus Method in Nonlinear Settlement Analysis of Sand Foundation Application de la méthode du module tangent dans le calcul du tassement non-linéaire de fondations sur sol sableux Yang G.-H., Luo Y.-D., Zhang Y.-C., Wang E.-Q.

3483

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Committees / Comités Conference Advisory Committee (ICSMGE) / Comité consultatif (SIMSG) President /Président Jean-Louis Briaud (ISSMGE President) Members / Membres Ivan Vanicek (Vice President Europe), Neil Taylor (General Secretary), Mamdouh Hamza (17ICSMGE Chairman), Pedro Seco e Pinto (ISSMGE Immediate Past President), Suzanne Lacasse (President of the Technical Oversight Committee), Roger Frank (18ICSMGE President of the Strategic Advisory Committee), Philippe Mestat (18thCSMGE President of the Conference Organizing Committee), Alain Guilloux (18thCSMGE Vice-President of the Conference Organizing Committee), Conference Organizing Committee / Comité d’organisation President / Président Philippe Mestat Vice-President / Vice-président Alain Guilloux Members / Membres Séverine Beaunier, Valérie Bernhardt, Nathalie Borie, Laurent Briançon, Yu-Jun Cui, Pierre Delage, Jacques Desrues, François Depardon, Philippe Gotteland, Pierre Habib, Roger Frank, Philippe Liausu, Stéphane Monleau, Claude Plumelle, Alain Puech, Jacques Robert, Frédéric Rocher-Lacoste, David Remaud, Françoise Ropers, François Schlosser President / Président Roger Frank

Strategic Advisory Committee / Comité stratégique consultatif

Vice-president / Vice-président Alain Puech President / Président François Schlosser

Scientific Committee / Commission scientifique

Vice-presidents / Vice-présidents Jacques Desrues, Pierre Delage Members / Membres Dietmar Adams, Dominique Allagnat, Eduardo Alonso, Gabriel Auvinet, Mounir Bouassida, Malek Bouazza, Sylvie Bretelle, Denys Breysse, Bernardo Caicedo, Robert Charlier, Christophe Chevalier, Alain Corfdir, Michael Davies, Peter Day, Hervé Di Benedetto, Claudio Di Prisco, Rich Finno, Etienne Flavigny, Bertrand, François, Roger Frank, Jean-Jacques Fry, Christophe Gaudin, Antonio Gens, Antonio Gomez-Correia, Roland, Gourvès, Yves Guerpillon, Pierre Yves Hicher, Robert Holtz, Catherine Jacquard, Richard Jardine, Richard, Kastner, Jean-Bernard Kazmierczak, Zoubeir Lafhaj, Serge Lambert, Eric Leber, Serge Leroueil, Michael Lisyuk, Juan Martinez, Farimah Masrouri, Hormoz Modaressi, Arezou Modaressi, Franz Molenkamp, David Muir Wood, Charles Ng, Fusao Oka, Trevor Orr, Olivier Pal, Anne Pantet, Manolo Pastor, Alain Puech, Françoise Ropers, Pierre Schmitt, Bruno Simon, Claudio Tamagnini, Jean-François Thimus , Luc Thorel, Christos Tsatsanifos, Serge Varaksin, Laurent Vulliet The Scientific Committee acknowledges the contribution of  the following persons to the edition of the Proceeding/ Le comité scientifique reconnaît la contribution à la réalisation des actes des personnes suivantes : P. Bésuelle, M. Boulon, S. Burlon, J. Canou, C. Chevalier, G. Combe, J.-C. Dupla, S. Hemmati, O. Jenck, A. Le Kouby, O. Plé, A. Pouya, S. Salager, J. Sulem

Committee for the French-speaking World / Commission pour la francophonie President / Président Jean-Pierre Magnan Vice-presidents / Vice-présidents Claude Plumelle , Mounir Bouassida Committee for Sponsors and Exhibition / Commission pour les sponsors et l’exposition President / Président Valérie Bernhardt Vice-president / Vice-président Jacques Robert Organizing Committee for the 5th Young Geotechnical Engineers Conference (5th iYGEC)/ Comité d’organisation du Congrès des jeunes géotechniciens (CIJG) President / Président Yu-Jun Cui Vice-president / Vice-président Fabrice Emeriault Members /Membres Fhad Cuira, Siavash Ghabezloo, Jean-Michel Pereira, Hugo Ravel, Michael Reboul, Anh Minh Tang, Séverine Beaunier

L

Foreword The French Society for Soil Mechanics and Geotechnical Engineering (CFMS) is most happy to host the 18th International Conference on Soil Mechanics and Geotechnical Engineering (18th ICSMGE) in Paris, France, from Monday 2 to Friday 6 September 2013. The main theme of the Conference is “Challenges and Innovations in Geotechnics”. In agreement with ISSMGE vision for strengthening the role of the Technical Committees (TCs), the 18th ICSMGE Paris 2013 adopted a new format. The two first days are devoted to plenary sessions with the Terzaghi Oration, seven ISSMGE Honour lectures proposed by the TCs and three Special lectures proposed by CFMS. The two following days are devoted to parallel sessions organised by the TCs: they include 28 Discussion Sessions and 22 Workshops. Whereas Workshops have a free format, the Discussion Sessions are meant to discuss the papers accepted by the ISSMGE Member Societies and presented in the four volumes of these Proceedings. The structure of the Proceedings corresponds to the organisation of the Conference. They start with the Terzaghi Oration, the Honour lectures and the Special lectures. Then, the papers are presented according to the relevant responsible TC. They are introduced by a TC General Report The Proceedings also include the papers on Shallow Foundations (Session and General Report organised by CFMS, as no TC covers this subject), on Dams, Dykes and Levees (organised jointly by TC 201 and TC 210), on Geothermal issues (organised jointly by TC 212 and TC 307) and finally the papers on Historic sites, as well as on some case studies (organised by TC 301). To enhance the diffusion of knowledge feee of charge, no transfer of copyright was requested from the authors of the papers published in these volumes. All the papers, together with late contributions, will be made available free of charge on the various appropriate Internet websites. It was quite a fruitful and exciting experience for the Scientific Committee to work hand in hand with the TCs for the organisation of this Conference. They were enthusiastic and efficient. The Scientific Committee is most grateful to the General Reporters, Chairs, Vice-Chairs and Secretaries of the TCs for their great help in making the 18th ICSMGE in Paris a most successful scientific and technical event.

Pierre Delage, Jacques Desrues, Roger Frank, Alain Puech, François Schlosser

1

Avant-propos Le Comité français de mécanique des sols et de géotechnique (CFMS) a le grand plaisir d’accueillir le 18e Congrès international de mécanique des sols et de géotechnique (CIMSG) à Paris, du lundi 2 au vendredi 6 septembre 2013. Le congrès est organisé autour du thème principal : « Défis et Innovations en Géotechnique ». En cohérence avec le souhait de la Société internationale (SIMSG) de renforcer le rôle des comités techniques (CTs), le 18e CIMSG Paris 2013 adopte le nouveau format suivant : les deux premiers jours sont consacrés aux sessions plénières avec l’allocution Terzaghi, sept conférences honorifiques de la SIMSG, proposées par les CTs, et trois conférences spéciales proposées par le CFMS. Les deux jours suivants sont consacrés aux sessions parallèles organisées par les CTs, comprenant 28 sessions de discussion et 22 ateliers. Les sessions de discussion sont le lieu du débat autour des contributions acceptées par les sociétés membres de la SIMSG et réunies dans ces actes en 4 tomes du congrès. Le format des ateliers est laissé à la discrétion de leurs organisateurs. La structure des actes correspond à celle du congrès : ils commencent par l’allocution Terzaghi, les conférences honorifiques, puis les conférences spéciales. Viennent ensuite les contributions, réunies par CT, et précédées du rapport général du CT. Les actes comportent également les contributions sur les fondations superficielles (session et rapport général organisés par le CFMS, car ce sujet n’est celui d’aucun CT), sur les barrages, les digues et les levées (organisation conjointe par les CTs 201 et 210), sur la géothermie (par les CTs 212 et 307), et enfin sur les sites historiques ainsi qu’un certain nombre d’études de cas (par le CT 301). Dans le but de faciliter la circulation des connaissances, il n’a pas été demandé aux auteurs de transférer leurs droits pour les contributions publiées dans les actes. Toutes ces contributions, ainsi que des contributions disponibles après le congrès, seront mises en accès gratuit sur divers sites internet appropriés. La collaboration étroite avec les CTs pour l’organisation de ce congrès a été pour la commission scientifique une expérience riche et passionnante, grâce à l’enthousiasme et à l’efficacité de ses interlocuteurs. La commission scientifique remercie vivement les rapporteurs généraux, les présidents, les vice-présidents et les secrétaires des CTs pour leur aide décisive en vue de faire de ce 18e CIMSG un événement scientifique et technique des plus réussis.

Pierre Delage, Jacques Desrues, Roger Frank, Alain Puech, François Schlosser

2

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

ISSMGE – The State of the Society (2009-2013) SIMSG – État de la Société (2009-2013) Briaud J.-L. President of ISSMGE, Professor and Holder of the Buchanan Chair, Zachry Dpt. Of Civil Engineering, Texas A&M University, College Station, Texas, 77843-3136, USA. [email protected]

Distinguished Colleagues, Dear Friends,

1

The very first thing I wish to tell you is thank you, thank you for letting me serve you as your President for the last four years. You have given me four of the very best and most exciting years of my professional career. It has been an honor and a true pleasure for me to work with everyone of you for the betterment of our profession. Sometime people ask me how I feel about the Presidency, I answer it feels like a very hard working vacation! You elected me in Alexandria, Egypt in 2009 and I suddenly found myself on a list next to the names of Terzaghi, Peck, Cassagrande, Skempton, Kerisel, and many other giants of our field (Fig. 1). This prestigious and enviable position also placed a tremendous sense of responsibility on my shoulders and generated a lot of pressure for me to do the very best job I could do. I can assure you that I gave it my very best effort, at the detriment of some of my other responsibilities in life. My wife Janet kept me honest during all this time. I recall asking her how she felt to be married to the President of the International Society. She promptly answered President Briaud don’t forget to take care of the garbage!!

My vision as President was a. To involve the membership and generate a sense of ownership in every one of you. I wanted you to feel that you were part of your professional family and that the family cared about you. This would be done for example by creating Board Level Committees where more members could participate and make high level decisions, by writing progress report to ensure that you felt connected, and by creating new awards to recognize those who excel in our profession.. b. To modernize the society and further advance it into the electronic age. This would be done for example by starting a series of free webinars, revamping the web site, creating GeoWorld, transferring the Lexicon to an addressable data base available on the web site, having the Board start meeting by Skype conference calls to save money. c. To help developing countries and the young geotechnical engineers. This would be done for example by raising money for the new ISSMGE Foundation which would receive applications and distribute grants, by creating a special group with direct access to the President. d. To mobilize more actively the practitioners side of our society and help bridge the gap between academics and practitioners. This would be done by creating a special group for practitioners with direct access to the president and recruiting more Corporate Associates into the Society. e. To enhance the image of the geotechnical engineer worldwide. This would be advanced by creating a Public Relations Group dedicated to simple steps that would increase the visibility of our profession. My basic tactic to realize my vision was pretty simple: 1. Develop a vision of what I wanted to accomplish 2. Surround myself with very smart people. Here I was very lucky to be able to convince the outstanding people including Harry Poulos, Suzanne Lacasse, Mike Jamiolkowski, Marc Ballouz, Dimitris Zekkos, François Schlosser, Jennifer Nicks, Michael Lisyuk. 3. Share with them my vision and check if they truly embraced it. 4. Give them a lot of freedom and support. 5. Be a strong cheer leader for those who did well 6. Be a gentle but steady nudge for those who dragged the team down 7. Keep thinking and acting with a vision for the relentless pursuit of excellence in a just and friendly atmosphere.

VISION

Fig. 1 Presidents of ISSMGE

Fig. 2 The 2009-2013 ISSMGE Board Members (in India)

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3. The tenure of the chairs of the TCs is four years renewable once. New chairs are suggested to TOC and the President by the members of the TC. 4. The TCs send a progress report to TOC every two years on which basis TOC decides to renew the TC or not but always after conferring with the President. 5. A short video was created by the public relations committee to explain in layman’s terms what geotechnical engineers do. 6. A number of innovations were created by IDC and are detailed subsequently.

THE BOARD (2009-2013)

I had a great team of 11 Board members who helped me accomplish all those initiatives. The Board members are shown in Fig. 2. Standing and from left to right are Samuel Ejezie (Vice President for Africa), Ikuo Towhata (Appointed board member), Ivan Vanicek (Vice President for Europe), Roger Frank (Appointed board member), Charles Ng (Appointed board member), Roberto Terzariol (Vice President for North America). Sitting and from left to right are Askar Zhussupbekov (Vice President for Asia), Michael Davies (Vice President for Australasia, first vice president and treasurer), Neil Taylor (Secretary General), Jean-Louis Briaud (President), Pedro Pinto (Past President), and Gabriel Auvinet (Vice President for North America). 3

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MEMBER SOCIETIES

We have a total of 86 member societies (Fig. 4). On the map of Fig. 4, the member societies are in dark. As you can see from the map, we need to continue our work in Africa to bring in more countries from that region to join ISSMGE. During the last four years two societies lost their membership because of repeated lack of dues payment but three new societies joined ISSMGE: Belarus, Chinese Taipei, and Lebanon. The total number of individual members increased from 18561 in 2009 to 19755 in 2013 or a 6.4 % increase. The members are distributed a follows: 1. Africa: 875 2. Asia: 3673 3. Australasia: 1590 4. Europe: 7985 5. North America: 4285 6. South America: 1347 The largest member societies are the USA (3294) followed by Japan (1155) and the UK (1130). The smallest society has 13 members. All societies have one vote.

BOARD LEVEL COMMITTEES

One of the first step was the creation of Board Level Committees (BLC) (Fig. 3) to engage more members in the affairs of ISSMGE. This process allowed me to have the participation of some 100 new people in charge of major decisions for The Society. The Technical Oversight Committee (TOC) chaired by Suzanne Lacasse in Norway was in charge of quality control for all 29 ISSMGE Technical Committees (TCs). The Membership, Practitioners, and Academicians Committee (MPAC) chaired by Harry Poulos in Australia was in charge of customer service for our 86 member societies including bringing academics and practitioners closer together. The Innovation and Development Committee (IDC) chaired by Dimitrios Zekkos in the USA was in charge of impacting The Society with new ideas and development of these ideas. In life, we rarely take the time to think so I decided that I would create a group whose job it would be to think. The Awards Committee (AWAC) chaired by Francois Schlosser in France would handle awards guidelines, awards decisions, and the creation of new awards if necessary. The Public Relations Committee (PRC) chaired by Marc Ballouz of Lebanon would start work on making geotechnical engineering more visible. The Students and Young Members Presidential Group (SYMPG) chaired by Jennifer Nicks in the USA would work directly with the President to accomplish some of the goals that would better serve that part of our Society. The Corporate Associates Presidential Group (CAPG) chaired by Michael Lisyuk would play a similar role for practitioners.

Fig. 4 ISSMGE Member Societies in 2013 ISSMGE Members and Member Societies

5 FedIGS Board Students and Young  Members Pres Group Jennifer Nicks (USA)

ISSMGE Council

ISSMGE Secretariat

ISSMGE Board

ISSMGE Foundation Harry Poulos (Australia)

Corporate Associates  Presidential Group Michael Lisyuk (Russia) Technical Oversight Committee Suzanne Lacasse (Norway)

We had a great discussion on the possible change of name of the society. The proposal was for ISSMGE to become ISGE: the International Society for Geotechnical Engineering. Arguments in favor and against were presented at the Council meeting in Toronto in 2011. The motion was proposed by several countries and the vote was 23 yes, 39 no, 1 abstain. We had a wonderful and professional discussion on this topic which brought out the passion all of us have for our profession. One of my goal during my presidency has been to engage the membership, I believe this topic definitely contributed to that. This was a very meaningful debate. It is my prediction that the name change to ISGE is only a matter of time but it may be a couple of decades before it occurs; soil mechanics is in our blood but it does not have to be in our name. I further predict that the word geotechnical engineering will soon become geo-engineering.

Awards Committee Francois Schlosser (France) Membership, Practitioners, and Academicians Committee Harry Poulos  (Australia)

Public Relations Committee Marc Ballouz (Lebanon)

THE NAME OF OUR SOCIETY

Innovations and Development Committee Dimitris Zekkos (USA)

Fig. 3 ISSMGE Organization Chart.

Some of the accomplishments and changes created by these Board Level Committees with subsequent approval of the Board are listed below 1. Young members can participate in Technical Committees as corresponding members without limit. They have to be nominated by the member society 2. Technical committees are no longer disbanded when a new President is elected. They continue right through the president election. However TOC and the President retain the right of closing a TC if it does not perform or change the leadership if the chair does not perform well.

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TECHNICAL COMMITTEES

Fig. 5 Location of the TC Chairs and sponsoring member societies

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The Technical Committees (TCs) were reorganized in three categories (Table 1), fundamental topics (7 TCs), applications (16 TCs), and impact on society (6 TCs), for a total of 29 TCs.

The location of the chairs and host society of the TCs is shown on Fig. 5.

Table 1 ISSMGE Technical Committees

Cat egor y Fun dam ental s

TC #

TC Official Name

Host Country

TC Chair

101 102 103

Laboratory Stress Strain Strength Testing of Geomaterials Ground Property Characterization from In-Situ Tests Numerical Methods in Geomechanics

104

Physical Modelling in Geotechnics

105 106 107

H. Di Benedetto P. Mayne K. T. Chau S. Springman (‘til 1 July 2010) C. Gaudin M. Bolton/M. Hyodo E. Alonso K. Ampadu

Netherlands

M. A. Van

Portugal

A. Gomes Correia

203

Geo-Mechanics from Micro to Macro Unsaturated Soils Laterites and Lateritic Soils Geotechnical Aspects of Dykes and Levees, Shore Protection and Land Reclamation Transportation Geotechnics Earthquake Geotechnical Engineering and Associated Problems

France USA Hong Kong Switzerland/ Australia UK/Japan Spain Ghana

Greece

K. Pitilakis

204

Underground Construction in Soft Ground

R. Kastner/A. Bezuijen

205 206 207 208 209 210 211 212 213 214 215 216 301 302 303 304 305

Limit State design in Geotechnical Engineering Interactive Geotechnical design Soil-Structure Interaction and Retaining Walls Slope Stability in Engineering Practice Offshore Geotechnics Dams and Embankments Ground Improvement Deep Foundations Geotechnics of Soil Erosion Foundation Engineering for Difficult Soft Soil Conditions Environmental Geotechnics Frost Geotechnics Preservation of Historic Sites Forensic Geotechnical Engineering Coastal and River Disaster Mitigation and Rehabilitation Engineering Practice of Risk Assessment and Management Geotechnical Infrastructure for Megacities and New Capitals Geo-Engineering Education (include aspects of software in use) Sustainability in Geotechnical Engineering

France/ Netherlands UK Canada Russia Canada USA China France Germany Germany Mexico Italy Norway Italy India Japan Singapore Brazil Australia

M. Jaksa

Canada

D. Basu

201 202

App licat ions

Imp act on soci ety

306 307

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B. Simpson K. Been V. Ulitsky J. Fannin P. Jeanjean Z. Xu S. Varaksin R. Katzenbach M. Heibaum J. L. Rangel M. Manassero A. Instanaes C. Viggiani V. V. S. Rao S. Iai K. K. Phoon A. Negro

the speaker through voice over IP and watch the slides on their computer screen. Fig. 7 shows the list of webinars offered over the last two years and the location of the computers connected worldwide for the first webinar. A contract was established with a web service company to facilitate the connection with many participants. The webinar series started in 2011 and the President gave the first webinar. Until August 2013 the webinars have been free and the recordings have been kept on the ISSMGE web site for free access. These recorded webinars have been accessed 1664 times since they have been uploaded three months ago. All speakers have been generous and offered to present the webinars for free as a gift to their fellow geotechnical engineers.

HONOR LECTURES

The TCs were given the opportunity to create an honour lecture named after one of the giants in their field. There were already 2 such lectures in 2009 (The Ishihara Lecture and the Mitchell Lecture), 7 more were created between 2009 and 2013 as listed on Fig. 6. Many of them were presented in Paris at the conference. Note that honour lectures are not necessarily permanent. They are created for eight years renewable by decision of the technical committee and approval of the Board.

ISHIHARA ‐ Earthquake MITCHELL – Site characterization BISHOP – Laboratory testing KERISEL – Monument preservation SCHOFIELD – Physical modeling McCLELLAND – Offshore geotechnics FUJITA – Underground construction MENARD – Soil Improvement ROWE – Environmental geotechnics Fig. 6 ISSMGE Honour Lectures

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WEBINARS

Webinars are lectures presented over the internet as follows. The speaker is at her or his desk in front of the computer screen. The speaker talks and advances the power point slides as would be done in a conference setting. The participants sit in front of their computer many kilometers away and listen to the voice of

Fig. 7 Webinar series and location of computers connected to the first ISSMGE webinar.

1. Scour and Erosion – Briaud, USA, 23rd Aug 2011

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2. Intelligent Compaction – Correia & Chang, Portugal, 25th Oct 2011 3. Eurocode- Bond, UK, 19th Dec 2011 4. Risk and Geotech Engrg – Medina & Uzielli, USA, 24th Feb 2012 5. Landfill liners – Rowe, Canada, April 2012 6. Unsaturated soils – Alonso, Spain, July 2012 7. Pile driving – Rausche, USA, September 2012 8. Earthquake engineering – Towhata, Japan, November 2012 9. Geosynthetics – Koerner, USA, January 2013 10. Ground Improvement – Varaksin/Huybrechts, Belgium, March 2013 11. Geophysics – Foti, Italy, May 2013 12. Foundations of very tall structures – Poulos, July 2013 9

SAN DIEGO, USA

DUBAI, UAE

CAIRO, EGYPT

MELBOURNE, AUSTRALIA

SAMARKAND, UZBEKHISTAN

PORT AU PRINCE, HAITI

LAGOS, NIGERIA

LANGZHOU, CHINA

AWARDS

In 2009, we had the Terzaghi Oration which is selected by the President of the Society alone, the Kevin Nash Gold Medal decided by the Council of Past Presidents, and three young geotechnical engineer awards decided by a committee of the Board. After calculating the ratio of awards offered by ISSMGE over the number of individual members of ISSMGE, I discovered that this ratio was extremely small compared to most other professional societies. We created 7 new awards as shown in Fig. 8. Then we created the Awards committee (AWAC) to finalize the awards descriptions, handle the collection of nominations and the selection process. The Board would make the final choice among the two candidates recommended by the Awards committee. The awards will be given at the Awards lunch in Paris and will be recorded on the ISSMGE web site. Terzaghi Oration Kevin Nash Gold Medal 3 Young Geotechnical Engineer Awards Outstanding Technical Committee Outstanding Member Society Outstanding Geotechnical Project Outstanding Innovator Outstanding Young Geotechnical Engineer Outstanding Public Relations Best paper in the Int. J. Geoeng. Case Hist. 9 Named Lectures

Fig. 8 ISSMGE Awards

10 TRAVEL I travelled extensively over the last 4 years with a total of 80 trips as shown in Fig. 9. During those trips I met so many people and made so many new friends. It was always a pleasure to meet geotechnical engineers throughout the world and I learned so much. I realized how much of a difference there is in the standard of living across the globe and that these differences cannot be solved by engineering and medicine alone. The biggest impediment to progress in some countries is corruption. Other impediments to an increase in the standard of living are lack of education and transportation. Until such basic problems are solved, the third world cannot rapidly improve. I kept many photographs of my trips and will continue to appreciate them as very special moments (Fig. 10).

Fig. 10 President Briaud on the road

11 THE ISSMGE FOUNDATION One of the realizations during my early travel was that there are huge inequalities in the salaries of geotechnical engineers throughout the world. Some people told me that their salary was $1000/year and added “How can I go to the conferences that you organize when the registration alone approaches one year salary”. This is when I decided to create the ISSMGE Foundation. By the way, it seemed very appropriate for a geotechnical engineering organization to have a Foundation! Harry Poulos agreed to look after its functioning and to head the grant distribution process. Today, any member of ISSMGE can apply for a grant from the Foundation. The application form and the rules are on our web site at http://www.issmge.org/en/issmge-foundation. Many geotechnical engineers, geotechnical companies, member societies, and even Technical Committees have contributed to the Foundation (Fig. 11) which currently has about $140,000 and has awarded grants to 19 people worldwide. Remember this saying that when you die, the only part of you that does not die with you is what you have given to others.

Fig. 9 The 80 places I visited during my Presidency.

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13 THE INTERNATIONAL JOURNAL OF GEOENGINEERING CASE HISTORIES Practitioners often complain that geotechnical journals are too academically oriented and that there is little useful to them. The IJGCS fills that gap: (http://www.issmge.org/en/resources/international-journal-ofgeoengineering-case-histories). Born a few years ago in the mind of Dimitris Zekkos, the IJGCS was endorsed by ISSMGE in early 2009 and has seen slow but steady growth. It is free of charge, on line, in color, with embedded spread sheet data when clicking on the figures. It is particularly welcome by developing countries which have access to high quality papers for free. It is not only useful to practitioners but also to professors who can use the case histories for their students in class. Jonathan Bray was the first editor in chief followed recently by Pedro Pinto. The ISSMGE TCs now have the opportunity of setting up special issues and the ISI rating is around the corner. The future of the IJGCS is very bright. I urge all of you to consider publishing a high quality case history in IJGCS. In life you have your financial wealth potential and your intellectual wealth potential. Publishing a case history in IJGCS is making an intellectual gift to developing countries: be generous and take the time to publish in IJGCS.

Fig. 11 ISSMGE 22 Foundation donors

12 CORPORATE ASSOCIATES ISSMGE Corporate Associates (Fig. 12) are geotechnical engineering companies, including consultants, contractors, and manufacturers who pay dues ($1500/year) to ISSMGE for a list of benefits (http://www.issmge.org/en/corporate-associates) and to support the profession. The Corporate Associates representatives (one per company) also form the Corporate Associate Presidential Group under the leadership of Michael Lisyuk of Russia. This group was created to work on aspects of ISSMGE which could benefit practitioners more specifically. In 2009 we had 21 CAs, today (2013) we have 43 CAs. This remarkable increase in the number of CAs is due to the hard work of many people and is very welcome. However this number still pales compared to the number of CAs in other international societies closely associated with ISSMGE who have more than 100 CAs. If you see your company logo on Fig. 12 we really appreciate your support. If you don’t, please consider joining and supporting your profession.

Fig. 13 GeoMap within GeoWorld: the new geotechnical engineers interaction medium

14 GEOWORLD Again born in the mind of Dimitris Zekkos and endorsed by IDC and SYMPG, GeoWorld (http://www.mygeoworld.info/) is to geotechnical engineers what Facebook is to social networking. It allows geotechnical engineers in the world to interact and make friends on line, to exchange questions and answers on various topics, to post examples, and to become even more connected internationally. Geoworld was launched in October 2011 and has now reached 2600 individual members, 160 companies, and 76 professional organizations. GeoMap is a new application within GeoWorld which allows you to find out members and companies in any geographic area by clicking on the GeoMap (Fig. 13). You can also find the location of upcoming conferences worldwide and the location of the case histories published in the IJGCS. 15 THE NEW ISSMGE WEB SITE Our new web site was launched in 2012. It was changed to allow ISSMGE to incorporate the latest technology and to modernize the look of the pages while maintaining flexibility of access and modification by the Secretary General’s office. The new site has a new conferences database, has increased functionality, hosts the recorded webinars, and promotes the integration with GeoWorld. The number of visitors has nearly double in the short time since it has been open going from 2200 visitors in June 2012 to 4000 in March2013 (Fig. 14). It also now hosts the new electronic version of the Lexicon.

Fig. 12 ISSMGE 42 Corporate Associates

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Fig. 14 Traffic on the new ISSMGE web site over the last 10 months. Fig. 15 The 2013 ICSMGE Paris conference organizing committee and its chair Philippe Mestat (center front row).

16 LEXICON The Lexicon was started around 1953 with the translation of geotechnical engineering terms in three languages: English, French, and German. This was very quickly recognized as a very valuable resource and had reached 8 languages by 1981 (5th edition). It had stayed that way until about 3 years ago when I asked Dimitris Zekkos and the Innovation and Development Committee (IDC) to transform the paper copy into an electronic and addressable Excel spread sheet and if at all possible increase the number of languages. We now have an eLexicon on our web site with 12 languages. Note that the eLexicon was a huge amount of work and is a great example of team work across country borders by many member societies and enabled by a platform developed by Geoengineer.org. The e-Lexicon includes a web-based application that allows users to query the database and find the translation of a total of 1590 geotechnical terms in 12 languages, specifically: English, French, Spanish, Turkish, Chinese (traditional and simple), German, Japanese, Portuguese, Russian, Persian (Farsi), and Finnish.

21 THE PROGRESS REPORTS Communication helps to solve problems and to generate a sense of belonging. This is why I started the monthly progress report in November 2009. The other reason was to tell you what was being accomplished. Since I got elected on 9Oct2009, my monthly report came on the 9th of each month. It required a tremendous discipline and dedication to not miss any of them but it provided a regular self evaluation of my work and our progress. 22 FUTURE OF GEOTECHNICAL ENGINEERING It is always very difficult to predict the future. A 20 year forecast is easier than a 100 year forecast and a 1000 year forecast is nearly impossible. Yet if we go back in history about 1000 years ago to the time of the Tower of Pisa, we then realize that designing a foundation for that Tower today would be a very simple exercise. Then we wonder by extrapolation what geotechnical engineering will be like in another 1000 years. Will we have? 1. complete non intrusive site investigation of the entire soil volume, 2. automated 4D computer generated design by voice recognition and based on a target risk, 3. tiny and easily installed instruments to monitor geotechnical structures, 4. unmanned robotic machines working at great depth, 5. significant development of the underground, 6. extension of projects into the sea, 7. soil structure interaction extended to thermal and magnetic engineering 8. failures down to a minimum, 9. expert systems to optimize repairs of defective geotechnical engineering projects, 10. geospace engineering of other planets, 11. geotechnical engineers with advanced engineering judgment taught in universities, 12. no more lawyers because of the drastic increase in projects reliability (Fig. 16).

17 THE ISSMGE BULLETIN The ISSMGE Bulletin was remarkably well handled by Ikuo Towhata as Editor in Chief and his team of editors. The Bulletin grew significantly in size and content under his leadership. Furthermore it went from 4 issues per year to 6 issues per year. We are very grateful to him for this enormous responsibility. 18 THE SECRETARIAT IN LONDON Neil Taylor was our Secretary General for the period and faced his responsibility with great poise. I could always count on Neil to tell me what the bylaws said. Paloma Peers was his assistant and continued to be a rock in a soil’s world. I also want to thank my assistant Theresa Taeger for being so reliable and dedicated to perfection. 19 THE FINANCES The finances of ISSMGE are in very good shape. The Members Societies dues have not changed during the last 4 years yet we have started new free programs for our members such as the webinars. Our budgets over the last 4 years have been approximately balanced and our reserves are healthy. This gives me a good occasion to thank the United States National Society and the Geo-Institute of ASCE for contributing to my yearly budget. 20 THE PARIS CONFERENCE The 18th International Conference on Soil Mechanics and Geotechnical Engineering will take place in Paris from 2 to 5 September 2013 and judging by the outstanding preparation, it will be a magnificent success. Our professional family will get together, to learn from the best, to exchange ideas and practices, all this in a classy, distinguished, yet relaxed and fun atmosphere. We are very grateful to our host: the French member society and its sponsors. Most of the members of the organizing committee are shown in Fig. 15 including Philippe Mestat, Chair of the Committee (center front row).

Fig. 16 Improved reliability of geotechnical projects (courtesy of George Nasr, Lebanon)

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23 A SUCCESSFUL CAREER A successful career is built on a series of demonstrated successes by an individual alone or as part of a team. In the performance of your job, remember when you make a decision of any sort that one mistake will take ten successes to erase the mistake from the mind of your peers. This is why it is always important to concentrate and plan. Also remember before a challenging moment that you may have been through similar tough moment before and have done well; this reasoning will give you added confidence and lower the stress. The following are some thoughts on what is important in a career. They have been inspired from discussion with many engineers over time including Clyde Baker and personal experiences as well. 10. Chose the relentless pursuit of excellence as a way of life 9. Be curious. The discovery process is a fountain of youth 8. Work hard but balance your interests (fun, family, sport, art, world news) 7. Make lots of friends. Nurture your public relations 6. Look for solutions and not who is to blame. Leave that to the judge. 5. Be firm in your decisions but always fair and polite 4. Treat others as you wish to be treated, you will lead by example 3. Communication is the best way to solve problems. Convince through logic and data 2. Surround yourself with smart people and role models 1. Go after your dreams with vision and perseverance 24 GEOTECHNICAL ENGINEERING FOR THE PEOPLE, BY THE PEOPLE, WITH THE PEOPLE While we continue to advance the profession, there is also no doubt that we do not get the recognition that we deserve. If you go in the street today and say to a passerby “my child is a heart surgeon”, that person will be very impressed. If you then say my other child is a geotechnical engineer, you will likely be asked: “what it that?”. There is a need to enhance the public’s recognition and awareness of our profession and this is why we have created the Public Relations Committee led by Marc Ballouz. It will be a very long road before we are recognized as heart surgeons are but the only way we can make a real difference is if every one of you takes the time to explain it to the people in the streets. One of our best ambassadors is Ikuo Towhata from Japan who came up with this saying: “Geotechnical engineering for the people, by the people, and with the people”.

UZBEKISTAN

AUSTRALIA

MOZAMBIQUE

LEBANON

ROMANIA

BRAZIL

SPAIN

HUNGARY

EGYPT

VIETNAM

RUSSIA

ITALY

25 CONCLUSION If someone asked me what has been the most rewarding part of my presidency I would not hesitate and say that it is making so many new friends all over the world (Fig. 17). Bill Gates, the richest man in the world today, was asked “how do you measure success in life?”. I believe he responded something like: “by how many friends you have”. All of you have been very kind to me over the last 4 years. I do not know if I will ever be able to repay such kindness before I die but I can assure you that it did not go unnoticed and it was extremely appreciated. Everywhere I went it felt like coming home for a special event, you welcomed me in your daily life as if I were coming to see the family. You treated me like a close friend and made me feel comfortable. I believe in team work and the ultimate team is the family (Fig. 18). I think that we have developed a better sense of family in our society and we are stronger for it. I say good bye as your President, but it will be my pleasure to become again a regular member of ISSMGE and to continue to serve you to the best of my ability. You certainly can continue to count on me if I can help. While I will no longer be your president, I will have the same desire to help you and to help the professional family. You mean a lot to me. Thank you again for all your kindness, take care, and remember that happiness is a choice.

TEXAS A&M UNIVERSITY Fig. 17 So many new friends!

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To finish I will borrow a saying from ASFE. When it comes to the soil, when it comes to the Earth, you are the best. Indeed, you are the best people on Earth.

Fig. 18 The Professional family

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8th Terzaghi Oration Protecting society from landslides – the role of the geotechnical engineer 8e allocution Terzaghi La gestion de l’aléa glissements de terrain et le rôle de l’ingénieur géotechnicien Lacasse S. Norwegian Geotechnical Institute (NGI), Norway

ABSTRACT: Protecting society from landslides and reducing exposure and risk to population and property are areas where the geotechnical profession can practice both the art and the science of engineering legated by Karl Terzaghi. The paper presents several case studies of slope failure and examples of landslide risk management. Since factor of safety remains the practice’s main indicator to ensure slope safety, the significance of factor of safety is discussed. The geotechnical engineer’s role is not only to act as technologist providing judgment on factors of safety. The role has evolved to providing input in the evaluation of hazard, vulnerability and risk associated with landslides. The geotechnical profession should be increasingly perceived as reducing risk and protecting people. RÉSUMÉ : La réduction de l’aléa dû aux glissements de terrains est devenue l’une des sphères où l’ingénieur géotechnicien peut pratiquer l’art et la science que nous a légués Karl Terzaghi. L’article présente plusieurs études de cas de glissements et des exemples de gestion du risque au glissement. Puisque le facteur de sécurité demeure l’indicateur principal de la stabilité des pentes, l’article discute les implications du facteur de sécurité. Le rôle de l’ingénieur géotechnicien n’est plus simplement d’offrir un jugement sur le facteur de sécurité, mais aussi de générer les paramètres et l’analyse pour l’évaluation des aléas, de la vulnérabilité et des risques associés aux glissements de terrain. Notre profession devrait de plus en plus être perçue comme réduisant le risque et protégeant la société. KEYWORDS: landslide, slope stability, strain-softening, factor of safety, case studies, hazard, risk 1

Protecting society from landslide hazard and mitigating the exposure and risk to population and property is one of the issues where we can practice both the art and science legated to us by Karl Terzaghi. Landslide issues and how to protect population has become a key to recruiting concerned young talents to the geo-profession. This is the reason why the topic of landslides, in terms of protecting society, was selected for the 2013 Terzaghi Oration. The mandate of the Terzaghi Oration is to cover case histories derived from professional activities, and if possible to illustrate the dynamic interaction among consulting work, teaching, research and publication. The case studies selected for this Terzaghi Oration attempt to exemplify Karl Terzaghi's intellectual approach to engineering and geology. Landslides and the protection of society from its hazards are a well-suited topic to meet this mandate, as landslides require a good understanding of the geology and soil behaviour, and have ample room for improvement. The paper presents case studies of landslides and examples of landslide risk management. Since factor of safety remains the main indicator to ensure the safety of populations in practice, the significance of factor of safety is discussed. The role of the geotechnical engineer in protecting people is focused on as part of the conclusions.

INTRODUCTION

The ISSMGE hosted seven Terzaghi Orations. Table 1 lists the topics covered earlier. The topics reflect an evolution and a cross-section of our professional practice. The 1st Terzaghi Oration was on the progress over 30 years in the prediction of cliff side instability. The 2nd described the design of the giant offshore structures marking the start of the suction anchor concept now widely used around the world. The 3rd Oration looked into prediction and performance for embankments on soft clay and pile foundations. The Terzaghi Oration then gave us remarkable case studies, (1) the shattering Kobe earthquake in Japan and (2) how the movements of the Pisa tower can be curbed to preserve the tower for future generations. The importance of the interaction of soil and water for the Netherlands came with the 6th Oration. The 7th Terzaghi Oration marked the emergence of slender high-rise buildings and introduced us to their challenging foundations subjected to large vertical, lateral and moment loads. Table 1. Terzaghi Orations 1985-2009. Year Author Title 1985 T.W. Lambe Amuay landslides. Foundation engineering for the 1989 K. Høeg Gullfaks C offshore gravity structure. 1994 V. De Mello Revisiting our origins. Geotechnical aspects of the 1995 1997 K. Ishihara Kobe earthquake. Leaning tower of Pisa: End of an 2001 M. Jamiolkowski Odyssey. 2005 F. Barends Associating with advancing insight. Tall buildings and deep foundations – 2009 H.G. Poulos Middle East challenges.

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LANDSLIDE HAZARDS

Landslides represent a major threat to human life, constructed facilities, infrastructure and natural environment in many regions of the world. During the decade 2000-2009, natural disasters caused nearly one million fatalities, affecting nearly 2.5 billion people across the globe. In 2010 alone, 295,000 fatalities due to natural disasters were recorded by Munich RE (2011) and the overall economic losses were more than double those of 2009, for approximately the same number of natural catastrophes. Table 2 presents the 2010 natural catastrophe data published by Munich RE NatCat-

Over the past decade, the geotechnical profession has moved in a direction of increased awareness of both its role and contribution to a safer society, and the need for targeted communication has emerged more strongly than earlier.

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though belonging to “after-the-fact” sagacity, lessons learned will be especially focused upon. The following case studies are included (section number is given in parenthesis):  The Vestfossen slide in sensitive clay, Norway (4)  The Kattmarka slide triggered by blasting, Norway (5)  The Saint-Jude natural slope failure, Québec, Canada (6)  Recurrent sliding on Cap Lopez, Gabon (7)  The Ashcroft Thompson River landslides, BC, Canada (8)  The Aalesund slide, Norway (9)  The Storegga slide, NE Atlantic Continental margin (10). The following landslide risk management examples are also briefly presented:  Landslide prevention in Norway.  The SafeLand Project.  Slope safety in Hong Kong.  Preparedness.  A few recent developments.

SERVICE. Most of the increase is due to the increase in the exposed population. However, many lives could have been saved if more had been known about the risks associated with natural disasters and risk mitigation measures had been implemented. Urban development, increased infrastructure and rapid population rise contribute to increasing the vulnerability of humans and property to landslides. While earthquakes, floods, tsunamis and storms receive wide attention in the news, landslides are not recorded as a separate hazard by Munich Re. The European statistics from the past 100 years in Table 3 give the social-economic impact of landslides in Europe in the 20th century. The landslide frequency of about 20 major events per year in Europe is the highest compared to floods, earthquakes and cyclones. However, the number of fatalities and the quantity of material damage is far greater for earthquakes. Landslides are also frequently triggered by floods and earthquakes and are not statistically recorded as landslides, but as floods and earthquakes in the disaster databases. Tragically, developing countries are more severely affected by natural disasters than developed countries, especially in terms of lives lost (UNDP 2004, UNISDR 2009 and IFRC 2004). Table 4 shows the data compiled by IFRC (2001) for the decade 1991-2000. Of the total fatalities due to natural disasters, the highly developed countries accounted for 5 % of the casualties. In absolute numbers, the material damage and economic loss due to natural hazards in highly developed countries by far exceed those in developing nations. However, this reflects the grossly disproportionate values of fixed assets, rather than actual economic vulnerability. Table 2. Natural catastrophes in 2010 (Munich Re 2011) Events and Average 2010 2009 losses(MUSD) 2000-2009 No. of events 950 900 785 Overall losses 130,000 60,000 110,000 Insured losses 37,000 22,000 35,000 No. fatalities 295,000 11,000 77,000

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Average 1980-2009 615 95,000 23,000 66,000

THE VESTFOSSEN SLIDE Description of the landslide

The slide occurred in 1984 and involved 50,000 m3 of soil that propagated about 100 m in almost horizontal terrain until it stopped on the opposite side of the Vestfossen River, close to Drammen in Norway. The geometry before and after failure in Figure 1 shows the critical circular slip surface in the middle and other slip surfaces studies. The failure had a 150-m long run-out across the Vestfossen River, as illustrated at the top of Figure 1. The failure was triggered by a fill placed mid-slope when a new soccer stadium was to be built. During project planning, the slope was probably assumed to have sufficient safety margin because the new slope was not steeper than the original slope.

Table 3. Impact of natural disasters in Europe (1900-2000) Disaster Lose of life Material damage 45 floods 10,000 105 B€ 1700 landslides 16,000 200 B€ 32 earthquakes 239,000 325 B€ Table 4. Natural disasters between 1991 and 2000 (IFRC 2001). Countries No. of disasters No. of lives lost Low & medium dev. Countries 1838 649,400 Highly developed countries 719 16,200

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OVERVIEW OF CASE STUDIES

Professor Ralph B. Peck, Karl Terzaghi’s closest colleague, relied heavily on case studies to learn from and to develop innovative solutions. After Karl Terzaghi himself, no one has influenced our practice as strongly as Ralph B. Peck with his 65 years of practice. Ralph Peck had a philosophy of simplicity of communication, whereby “if you cannot reduce the presentation of a difficult engineering problem to just one sheet of paper, you will probably never understand it” (Course CE484, University of Illinois; DiBiagio 2013). While achieving one-page summaries for each case study was not possible in this Oration, an attempt was made to stick to Ralph B. Peck’s philosophy. Each case study is organized contains essentially four components: 1. Description of the landslide 2. Soil parameters 3. Analysis of the landslide 4. Lessons learned The summaries do not contain all the details for each case study. However, the details may be found in the references cited. Al-

Figure 1. Cross-section before and after the Vestfossen slide also showing the undrained shear strength from field vane tests.

4.2

Soil parameters

Below the drying crust, the clay had water content of 45% at depths 4 to 10 m. The water content decreased to 30% below 12 m. Laboratory fall cone tests indicated a clay with extremely high sensitivity with St ≈ 150-200 in the top 12 m, and St ≈ 50-

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the sum of driving forces. The calculations considered strain compatibility (Grimstad and Jostad 2012). The strain compatibility was achieved by finding the highest safety factor on a given slip surface for different constant shear deformations. Thereafter, the slip surface giving the lowest safety factor was located. The strain-compatible critical slip surface was not necessarily the same as for the case without strain compatibility. To do strain-compatible calculations, an idealized material model was used, as shown in Figure 4. The peak shear stress was taken at a shear strain of 1% in triaxial compression, 5% in direct simple shear and 10% in triaxial extension.

100 below 12 m. The overconsolidation ratio below the drying crust was 1.1, due to aging) Figure 1 provides profiles of undisturbed and remoulded undrained shear strength from the field vane test (FV). Figure 2 presents the undrained shear strength normalized with the effective overburden stress, p'o, from triaxial compression, direct simple shear and triaxial extension tests vs the inverse of the overconsolidation ratio (OCR). Specimens from depths of 7, 13 and 17 m were tested. Figure 3 illustrates three stress strain curves and effective stress paths from anisotropically consolidated triaxial compression tests. The residual shear strength and the peak shear strength for a “perfect” sample are also indicated with the dashed line. To simulate a “perfect” sample, the effective stress path of a perfect specimen follows an angle of 1:3 up to the failure line (Berre et al 2007).

Figure 4. Idealized anisotropic stress-strain model for straincompatibility modelling (Grimstad and Jostad 2012).

Figure 5 presents the results of the limit equilibrium stability analyses when the peak undrained shear strengths were used. The factors of safety obtained are listed in Table 5.

Figure 2. Normalized undrained shear strength, Vestfossen clay (Grimstad and Jostad, 2011a).

Table 5. Result of limiting equilibrium analyses of Vestfossen slide. Case Strain compatibility Factor of (Slip surface) safety Fill added No 1.01 (Fig. 5, top) Yes 0.93 Before addition of fill No 1.26 (Fig. 5, bottom) Yes 1.19

Figure 5. Result of limiting equilibrium analyses of Vestfossen slide (Grimstad and Jostad 2012).

Figure 3. Stress-strain curves and effective stress paths from triaxial compression tests, Vestfossen clay (Grimstad and Jostad, 2011b).

4.3

Including the strain compatibility criterion decreased the safety factor by about 7%. With the strain-compatible model and the added fill, the slip surface extended further beyond the toe. The safety factor of 1.2 for the case “before the addition of the fill”

Analyses of the slide

4.3.1 Limit equilibrium analyses The classic Fellenius method was used, where the factor of safety, FS, is calculated from the ratio of the sum of resisting to

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

was too optimistic, because the peak shear strengths were used and side shear was not included in the analyses. 4.3.2 Finite element analyses The material model NGI-ADPSoft (Grimstad and Jostad 2012) was used to model the sensitive clay. The model is a userdefined special version of the NGI-ADP model (Grimstad et al 2010; 2011) which was implemented as a standard material model into Plaxis (www.Plaxis.nl). The model is an elastoplastic model that describes the anisotropic behaviour of clays during undrained shear and includes post peak strain-softening. The model is suitable for modelling the initiation of progressive failure in sensitive clays under undrained loading. The model uses as input the data from anisotropically consolidated undrained triaxial compression (CAUC) tests, constant volume direct simple shear (DSS) tests and undrained triaxial extension (CAUE) tests. The input parameters are the peak undrained shear strength sup and the residual shear strength sur and the corresponding shear strains p and r along the shear stress-shear strain curves (Fig. 6). The curves start at an initial shear stress o with a slope equal to the initial shear modulus Go. In the calculations, Go is set equal to Gur. Plane strain compression and extension were assumed to be equal to the results of triaxial compression and extension tests. Through interpolation between the three curves, the model describes the general 3D anisotropic behaviour of the clay that depends on the actual orientation of the maximum shear deformation.

remaining part of the strain-softening curve towards residual governs post-failure displacements. The safety factor obtained by the finite element analysis before failure, without strain-softening and without strain compatibility was 1.28, which is very close to the 1.26 in Table 5. With the addition of fill, the safety factor from the finite element analyses was 1.0. Figure 8 illustrates the failure zone for the case of no strain-softening. The failure zone extends much further up slope and less at the toe than in the case with strainsoftening. The uncertainties in the analyses were mainly related to the strength in the drying crust, the initial effective stresses under the fill, and the thickness of the shear band after mobilization has been initiated.

Figure 6. NGI-ADPSoft model parameters (Fornes and Jostad, 2013).

The softening behaviour is governed by introducing a “nonlocal plastic shear strain”. The so-called “non-local strain” (Eringen, 1981) means that the plastic strain in a stress point (Gaussian integration point) is replaced by an integrated weighted average plastic strain within a specified zone around the point. The plastic strain and ensuing reduction in shear strength during softening become mesh independent, and are controlled by the input parameters. The shear band thickness and resulting brittleness are then also controlled by the input data (Grimstad and Jostad, 2011; Grimstad and Jostad 2012). Figure 7 illustrates the progressive development of the failure. Each diagram gives a snapshot for increasing incremental displacements (from NINC =40 to 160). The figure shows that it is possible to model strain-softening. The analysis did not include the in situ variation in sensitivity of the quick clay (clay is much less sensitive upstream), and without the complete effect of the drying crust, which, if included, would have limited the shearing at the toe (which is unrealistically large in Figure 7). Jostad and Grimstad (2011) found that the critical strain at which progressive failure starts to develop is low, and not large enough to remould the clay. It is therefore only the initial part of the strain-softening curve that is of interest for capacity. The

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Figure 7. Incremental displacements in modell of Vestfossen progressive failure (NGI 2012).

Figure 8. Contours of total displacements, model without strainsoftening, scale in m (NGI 2012).

In summary, it was possible to obtain a factor of safety of 1 when using a best estimate of the soil parameters and the NGIADPSoft model, but the stress-strain curves used in the analyses had to account for the strain-softening observed in laboratory tests. The finite element modelling of the deformation under the embankment load led to a progressive development of the failure in a nearly horizontal terrain. The failure occurred along a circular slip surface (as shown), which gradually progressed as a circular surface towards the river. This was very close to the observed displacements after the failure in 1984.

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4.3.3 Sensitivity analyses Jostad et al 2013 did sensitivity analyses to quantify the required reduction in peak undrained shear strength, Fsoftening, for sensitive clays. Figure 9 illustrates the results obtained. The analyses were done with the PLAXIS finite element code with the NGI ADPSoft model. A total of 500 Monte Carlo simulations were done. The average required reduction of the peak undrained shear strength in triaxial compression, direct simple shear and triaxial extension was 9% (Fsoftening= 1.09). The values of Fsoftening ranged between 1.02 and 1.27. Although Figure 9 shows scatter, the effect of softening increases with decreasing load. The values of Fsoftening of 1.0 were cases where failure occurred in the drying crust. For 2.5 % of the simulations, Fsoftening was greater than 1.2, and for 12% of the simulations, Fsoftening was greater than 1.15. For stronger strain-softening clays, the factor Fsoftening was considerably lower that for the softer clays with low failure load.

analyses. One can either apply a reduction factor on the peak undrained shear strength from triaxial compression, direct simple shear and triaxial extension tests, or one can apply different factors on each test type, e.g. 15% on the triaxial compression strength, 10% on the direct simple shear strength and 5% on the triaxial extension strength. Based on Figure 9, one should consider establishing a reduction factor as a function of clay type (or strength), type of slip surface and perhaps slope inclination and clay sensitivity. With the knowledge available today, an average reduction factor between 1.10 and 1.15 may be reasonable. More research on this topic is underway. 5 5.1

LANDSLIDE IN KATTMARKA Description of the landslide

On March 13 2009, about midday, in Kattmarka near Namsos north of Trondheim in Norway, a slide occurred, moving about 500.000 m3 of material in a scar measured afterwards of about 100 m width by 300 m length. The slide destroyed a highway and damaged four permanent dwellings and 6 summer residences. Seven persons, who had been transported on the slide, were rescued unharmed by helicopter. Figures 10 and 11 illustrate the slide that occurred. Figure 12 illustrates the sequence of the movements (from 1 to 5) based on observations and eyewitness accounts. The slide (part in Fig. 12) started about ½ minute after the blasting of rock as part of highway construction nearly (Fig. 12). Part 2 slid 2 minutes later, thereafter Parts 3, 4, and 5. The sliding activity lasted between 6 and 10 minutes. The construction project nearby was a widening of the road into the mountainside, adding sidewalks, and upgrading of sewers and pavement by the Norwegian Public Road Administration.

Figure 9. Required reduction in peak undrained shear strength (Fsoftening) vs failure load for all sensitivity analyses (Jostad et al 2013).

4.4

Lessons learned

For brittle materials such as highly sensitive and quick clays, the strain-softening behaviour needs to be taken into account in the stability analyses (Jostad et al. 2013; Fornes and Jostad 2013). The brittle nature of the failure and the strain-softening are such that the peak strength measured in the laboratory cannot be used directly in limit equilibrium analyses. The stability of long slip surfaces in brittle and sensitive soils cannot be calculated by classical limit equilibrium methods. The calculated material coefficient will be overestimated for long slip surfaces to a greater degree than for local slip surfaces. Failure on long slip surfaces generally develops progressively in time and space. The shear strength along part of the slip surface reduces significantly, moving towards the remoulded shear strength, while other parts are still in the pre-peak, hardening regime. The peak shear strength is not representative for the shear resistance along the potential slip surfaces. Stability calculations in practice are usually done by limit equilibrium approaches that account for horizontal, vertical and moment equilibrium. As no commercial software that fully accounts for progressive failure is available today, limit equilibrium methods will continue to be used in practice. In the case of Vestfossen, one should note that it was necessary to reduce the peak shear strength by an average of 10%, if limit equilibrium analysis was used. The reduction accounts indirectly for strain compatibility and time effects. The initiation and progressive failure were captured well by a large deformation finite element analysis with PLAXIS 2D (Grimstad and Jostad 2011), using the NGI-ADPSoft material model: the safety factor was then 1.0. One needs to establish a reduction in the peak shear strength required to account for the strain-softening in limit equilibrium

Figure 10. Photograph of Kattmarka landslide (photo: L.A. Holme).

5.2

Soil parameters

The soil investigations post-landslide revealed the presence of sensitive clay with lenses of silt and sand. The clay thickness was between 10 and 20 m above bedrock. There is a thin layer of moraine above the bedrock. The soil consists of clay layers, some more silty than others, with thin sand lenses at irregular intervals. More than half of the clay was quick clay. The water content was above the liquid limit. The overconsolidation ratio in the clay below a drying crust about 2.5 m thick decreased from 2.5 at a depth of 3 m to 1.5 at a depth of 14 m. Figure 13 presents the undrained shear strength profile for the area. On the basis of the test results, the undrained shear strength selected for the stability analyses was selected as: suC = 15(kPa) + 2.0(kPa/m) · z(m)

(1)

where suC is the undrained shear strength in triaxial compression and z is the depth in meters. The undrained shear strength was

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

highly anisotropic with suDSS equal to 0.70 × suC and suE equal to 0.40 × suC, where suDSS is the undrained shear strength from direct simple shear tests, and suE is the undrained shear strength in triaxial extension. In Figure 13, the undrained shear strength values derived from the cone penetration test (CPTU) via the cone factors Nkt and Nu. An analysis was done of the uncertainties in the undrained shear strength, and it was concluded that exceeding the value of the suC used in the analyses was less than 10 or 15%. The analyses were done with the computer codes PLAXIS and GeoSuite Stability (Lacasse et al 2013). The NGI-ADP soil model for anisotropic clays was used. The two programs gave the same safety factors. The PLAXIS analyses were run with a plane strain approximation, with partial compensation of the 3D effects with a stabilizing side shear. To model the condition “After blasting, before sliding”, a zone of remoulded clay was included immediately at the rock-clay interface, which dimension of 8 m by 4 m was based on observations in situ (after the slide) and calculations of shear strains due to the blasting (Nordal et al 2009). Table 3 lists the resulting safety factors.

Figure 11. 3-D model of Kattmarka area before and after landslide (terrain model from laser scanning plate) (NVE 2009).

Figure 13. Undrained shear strength from laboratory and in situ tests and profile selected for stability analyses. Table 6. Factor of safety before sliding Zone (Fig. 11) 1 2 3

Stability condition Before blasting After blasting, before sliding Before blasting After blasting, before sliding Before blasting After blasting, before sliding

Factor of safety, FS 1.20 0.97 1.19 1.06 1.02 ~0.90

The delay of ½ minute between the blast and the initiation of the slide (visual observation) can be in part explained by rate effects, whereby the high frequency of the load caused an increase in the strength, but as the clay at the top of the slope became remoulded under the added load from the rock slipping and pushing in the clay, the clay towards the bottom of the slope could not support the added load. The overstressed area towards the bottom of the slope in the Zone 3 cross-section is illustrated in Figure 14. Figure 15 illustrates the vertical interface rock-clay in Zone 1, and the blasting that triggered the slide. The cross-section after the slide is also shown on the figure. The location of the blasting holes is only approximate on the figure, as it was difficult to reconstruct the exact locations in the aftermath of the slide. The blast shot the rock face out about 1 meter into the

Figure 12. Movement succession in Kattmarka (Nordal et al 2009).

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nicipality, the Norwegian Public Road Administration, the geotechnical consultant and the contractor building the road.

sensitive clay. Geological investigations of the rock also indicated the following (Fig.15): a nearly vertical weakness zone in the bedrock (Plan K); a fault at an angle of 36 º inclining towards the clay (Plan 3) acting as sliding plane for the rock under blasting; and other weakness planes in the rock mass contributing, with Plan 3, to pushing the rock face into the quick clay.

Figure 14. Results of stability analyses before blasting, cross-section in Zone 3, Kattmarka landslide (Nordal et al 2009).

Figure 16. Modeling of effect of blasting in clay sediments in Kattmarka (remoulded clay in red and yellow zones) (Nordal et al 2009).

The Namsos municipality introduced in 2003 the following regulation: before approval of building plans, geotechnical documentation shall confirm that the stability is acceptable and shall not be impaired. This was not done for the road project in 2009. No geotechnical investigation was carried out at the site before detailed planning. This was partly due to budget limitations. Although it is acceptable for stakeholders with considerable local experience to work on the basis of their wide knowledge in a region, the developer should have stopped the building activities to do site investigations when soft clay was found close to the road during the preparation for the blasting. The geotechnical consultant was hired to study the stability of the slopes in surrounding areas of the project and not in the areas of Kattmarka, and his work had been limited to 80 hours. The consultant had indicated the stability problem at the Kattmarka location, but the proposed actions were not followed up. The stability of the area of road construction was not analysed, although this is required by the NVE (2011) regulations. The Kattmarka landslide led to new regulations and an increased focus on existing regulations, including:  the control and mapping of the clay-rock interface when blasting in marginally stable areas;  the requirement for geotechnical investigations early in the project planning process; and  the necessity for hazard and vulnerability analyses for projects that can endanger life and property.

Figure 15. Cross-section in Zone 1 at the time of blasting, Kattmarka landslide (Nordal et al 2009).

Nordal et al 2009 did analyses of the shear strain () in the sensitive clay as the rock mass detached by the blasting penetrated the clay. Figure 16 illustrates one of the results. With the finite element mesh in the top part of the figure for the PLAXIS dynamic analysis, the blasting was modeled by a penetrating element with a maximum velocity of 10 m/s and a total displacement of 0.5 m into the clay. Equivalent linear properties were used in the clay for this calculation. The bottom cross-section in Figure 16 illustrates the shear strain contours. Liquefaction was believed to occur when the shear strain is greater than 3%. The slide was triggered by the blasting. The blasting moved the rock face and a block pushed outward into the clay with considerable force and velocity, causing the surrounding clay to liquefy. The unexpected movement of the rock face was a consequence of two unfavourable conditions: (1) the a priori unknown orientation of the rock-clay interface and (2) planes of weakness in the rock mass. The sensitive quick clays, however, had already before construction a marginal stability. The developer did not know of how critical the stability was. 5.3

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THE SAINT-JUDE LANDSLIDE Description of the landslide

In the evening of May10, 2010, a large landslide occurred in the municipality of Saint-Jude, northeast of Montréal, in Québec, Canada (Locat et al 2012). The landslide happened without warning on the right bank of the Salvail River, and tragically took four lives. The landslide swept away the road, aqueduct and power and telephone lines. Figure 17 presents a photograph of the landslide and the location of the bed of the Salvail River completely blocked by the landslide. The plain at the top of the natural slope before failure was at an elevation of 28m, and the slope inclination was between 12

Lessons learned

The slide had dramatic consequences, and it was just a matter of good odds that no lives were lost. Many parties were involved in the planning, design and building process: the Namsos mu-

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

since 1950 showed that erosion was active more or less steadily at the foot of the slope and that small landslides associated with erosion had occurred. The erosion seemed to have intensified over the past 15 years.

and 18º, and perhaps 20º in sections close to the toe. The height of the slope involved in the sliding was about 22 m. The slide area had a width of 275 m parallel to the watercourse and a length of 150 m normal to the watercourse. A total area of 54,000 m2 was affected by the landslide. The morphology of the slide was typical of a spread (Varnes, 1078). The debris were a succession of long slices of deformed and dislocated material oriented normal to the direction of movement. Some of the debris took the form of a triangular prism and reminded of horsts, and these were displaced horizontally only. The horsts were separated by slices of relatively undisturbed material, just like a block having dropped due to some sort of faulting, and were called grabens (Fig. 18). In the back part of the slide, some blocks coming from an upper failure surface were pushed upward by movement and overlapped the adjacent lower slices. The investigation of the landslide was carried out by the Geotechnical and Geological Department of the Ministry of Transportation of Québec (Locat et al 2011).

Table 7. Index properties, Saint-Jude slide (after Locat et al 2011). Depth w Ip IL Soil description (%) (m) (%) Drying crust, sandy, 0-3.8 24-78 silty from 2 m Clay, some silt traces 3.8-26 65 20-37 2.0-1.0 of sand 26-31 Silty clay 45-75 21-37 0.7-1.0 Silt, clayey, some 31-37 15-25 12-29 0.5-1.5 sand, trace gravel Silt, sandy, some clay, 37-42.6 13-18 trace gravel, v.dense >42.6 Shale and sandstone w water content Ip plasticity index IL liquidity index Table 8. Stress and strength characteristics, Saint-Jude slide location (after Locat et al 2011). Depth su p'c  Soil description (m) (kPa) (kPa) (kN/m3) 0-3.8 Drying crust, sandy, 50-165 250-400 18.6 silty from 2 m 3.8-26 Clay, some silt 25-65 100-260 16.0 traces of sand 26-31 Clay, sandy 50-107 180-310 16.8 31-37 Silt, clayey, some 40-150 19.3 sand, trace gravel 37-42.6 Silt, sandy, some 20.7 clay, trace gravel >42.6 Shale and sandstone su undrained shear strength from field vane (CPTU Nkt = 13.5) preconsolidation stress (OCR = 1.4 at El. +15 and 1.0 at El. 0) p'c  soil unit weight

Figure 17. Saint-Jude landslide and location of Salvail river (dashed line) (Locat et al 2011).

Figure 18. Saint-Jude landslide: illustration of the horsts and grabens after the slide (Locat et al 2011).

6.2

Soil parameters

The soils involved in the landslide are mainly marine clay from the former Champlain Sea. The clay was sensitive, of medium to firm consistency, and had sensitivity ranging from 30 to 80 and liquidity index decreasing form 2 to 1 with depth. There was artesian pressure of 10 m above the river level at the landslide site. Table 7 and 8 describe a typical soil profile. Figure 19 gives an example of the cone resistance measured at the site after the failure. The profiling enabled the determination of the location of the slip surface (Elevation +4 in Fig. 19). 6.3

Figure 19. Example of piezocone results in Saint-Jude deposit and indication of slip surface at Elevation +4 (Locat et al 2011).

Analysis of the landslide

The analysis of the available data revealed that landslides had occurred earlier along the Salvail River. Aerial photographs

The slide is believed to have occurred as follows. Along the centreline of the slide, the slip surface developed at depth and

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was sub-horizontal. The slip surface was detected Elevation 2.5 m below the Salvail river bed (originally at Elevation 6 m) for the first two-third of its length. The last third of the slip surface was at Elevation 15 m near the scarp. These two levels of the failure surface explain the overlapping of blocks in the central part of the scar. Figures 20 and 21 and Table 9 present the results of a few of the stability analyses. Only the most critical of the 1000’s of slip surfaces are shown. Analyses under drained and undrained conditions were run. The analyses were run with the SEEP/W and SLOPE/W code (GeoStudio 2007 verG7.17; GeoSlope International). For the case of drained conditions (Fig. 20), the computed factor of safety (FS) was 0.98 with the Bishop method and 1.03 with the Morgenstern-Price method. The lower part of Figure 20 shows the area of all the circles giving a factor of safety of 1.05 or less. All critical slip surfaces pass below the river bed, which agrees with the observations after the slide. For a failure surface extending significantly up slope (horizontal distance of 80 m in Fig. 21), the factor of safety was about 1.3, showing that this was not the triggering rupture mechanism. For undrained conditions (Fig. 21), using the observed slip surface, the safety factor was about 2.3. Table 9. Results of stability analyses of St-Jude landslide. Case Slip surface Method Drained Circular Bishop (Fig.54) Horizontal Morgenstern-Price Drained Circular Bishop Entire slope Horizontal Morgenstern-Price Undrained Circular Bishop (Fig.55) Horizontal Morgenstern-Price

6.4

Lessons learned

The high pore pressures in the clay below the river bed resulted in very low effective stresses, and therefore low resistance in the clay. The conventional analysis of the failure with circular slip surface gave a safety factor of unity under drained conditions, but could not explain the observed extent of the slide. Locat (2007) and Locat et al (2008) made similar observations. The trigger of the landslide is believed to have been of natural origin. The stability was impaired by (1) the high artesian pore pressure at the toe of the slope and (2) shoreline erosion, also at the toe of the slope. Since the meteorological conditions did not show any heavy rainfall at the time of the landslide, the trigger of the movement was probably the continuous erosion of the toe, which had aggravated over the last 15 years. The failure probably occurred in two stages, the first a rotation, thereafter a translation, as suggested by the post-failure grabens and the horsts (Fig. 18). The movement stopped when the debris accumulated on the other side of the river bank generated sufficient resisting forces to re-establish equilibrium. As mitigation in Saint-Jude on the Salvail River, the height of the natural slope was reduced to a maximum of 10 m, the river was moved about 60 m further away from the road and the debris were left in the landslide scar.

FS 0.98 1.03 3.1 3.3 2.2 2.3

7 7.1

RECURRENT SLIDING ON CAP LOPEZ Description of the landslide

On Cap Lopez in Gabon, a large underwater slide took place in July 1971. Twenty years later, the coastline had same topography as in July 1971. It was important to assess whether sliding would recur, as adequate safety was required for the oil terminal installations on land. Pointe Odden in Gabon is known for its rapidly changing coastline due to complex erosion and sedimentation patterns, enhanced by the presence of a deep submarine canyon on the west side. A 3,000,000 m³ slide took place on Cap Lopez’ north end, called Pointe Odden in July 1971 (Fig. 22). Figure 23 shows some of the coastline movements since 1911.

Figure 20. Results of limiting equilibrium analysis of St-Jude landslide under drained conditions: top: Bishop method, critical slip surface; bottom: All slip surfaces giving Factor of safety ≤ 1.05 (Locat et al 2011).

Figure 21. Results of limiting equilibrium analysis of St-Jude landslide under undrained conditions: Morgenstern-Price method, observed slip surface (Locat et al 2011).

Figure 22. Topography before and after 1971 slide (Lacasse and Boisard 1996)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

is a 10- m deep, 20-m wide zone of compacted sand, placed to stop the propagation of a slide. In 1979, dredging was carried out at the east side of Pointe Odden to remove sand down to 912 metres below sea level. The surface sliding however appeared to have a self-remediating effect. Elf opted to monitor whether this auto-regulation was sufficient to keep the coast stable in the future.

Figure 24. Underwater cross-sections before sliding (Lacasse and Boisard 1996)

7.2

The sand of Cap Lopez is a fine to medium coarse, mainly quartz, sand. The grain size distribution falls within the limits established in the literature for flow slide materials, and is similar to the grain size of sands which have experienced flow slides. Consolidatedundrained triaxial compression tests on sand sampled on-shore on Pointe Odden show that the sand in a loose state has a contractive behaviour with considerable strain-softening at low shear strains. Cone penetration tests suggested that layers with very low density or very low shear strength can be present.

Figure 23. Observations of Cap Lopez 1911-1971 (Lacasse and Boisard 1996).

7.3

Witnesses to the July 1971 slide reported that the slide started at 02:00 in the night and continued until noon the next day. Land extending about 310 m into the sea disappeared, moving about 1000 m away into deeper water. The slide had typical funnel and fan shape often associated with flow slides. Surface sliding of a 3 m thickness over most of the seabed east of Pointe Odden occurred between 1988 and 1989. The slopes before sliding had an inclination between 8 and 9°. Between 1989 and 1992, the coastline did not move significantly. In March 1992, a new large slide occurred. At its deepest, the 1992 slide was 10-12 m deep and extended 350 m in the east direction. The 1971 slide was 30 m deep and extended 1000 m out to sea. Figure 24 illustrates the cross-sections before sliding in the interval between the 70s and 90s. The observations of the coastline (Fig.23) suggest recurrent sliding, perhaps every 15 to 20 years as indicated in Table 8. Table 8. Periodicity of slides on Cap Lopez Approx. Date Event Slide (?) 1911-1920 Slide (?) 1930-1937 Slide in 1957(?) 1946-1957 Slide 1971 Slide 1992

Soil parameters

Analysis of the slide

Stability analyses considered both a drained situation and an undrained situation with development of excess pore pressures. The slides seem to have been triggered by a small increase of shear stress in a layer of looser/weaker material. The slide of March 1992 suggested that only a small additional amount of sand or a small change in pore water pressure was sufficient to trigger a large slide. On the basis of over 70 observations since 1971, the limiting inclination of the slope was 8-9°. Slopes less than 8° were always stable, slopes greater than 9° slid. Edgers and Karlsrud (1982) studied the mechanisms of submarine slide run-out with case studies. Figure 25 presents run-out distance as a function of the sliding volume. Observations for Cap Lopez are added to the graph, as well as a number of larger underwater slides that have been mapped in recent years (Canals et al 2004). The Cap Lopez slides plot at the limit of the underwater slides. The existing data show that (1) submarine slides may be triggered on very flat slopes; (2) the volume and run-out of submarine slides are by far greater than the volume of terrestrial slides; (3) the most predominant soil types with large run-out distances were fine sands and silts; (4) a trend for increasing relative run-out distance (run-out distance L over height drop H) with increasing slide volume. The Cap Lopez data follow this trend. A worst case scenario would be a deep-seated slide, extending close to the Quai des Chalands, and partly through the oil loading berth, causing environmental damage, affecting operation of the harbour and terminal, and requiring stabilisation of dangerous slopes after the slide. The "worst case" estimate was based on a series of “positive” and “negative” factors.

Time between events -15-20 years 15-20 years 15-205years 21 years

Elf Gabon took measures to protect the coastline. On the west side of the cape, protection walls were built every 25 m at frequent time intervals since the 50s to stop erosion and sand transport. From the Quai des Chalands to Pointe Odden on the east coast, a vibro-floated "wall" was built (Fig. 22). The "wall"

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7.4

On the one hand, among the «negative» factors: ‐ The slope between Pointe Odden and Quai des Chalands consists of young sediments with frequent and seasonal changes in the shoreline. ‐ The sediments south of the 1971 slide may be in a state of unstable equilibrium. The addition of a small quantity of sand can result in important surface sliding. ‐ The slopes of the seabed have an inclination prone to sliding; continuous weaker layers susceptible to slides may be present. ‐ The large surface slide in 1988-1989 did not preclude the possibility of the occurrence of a large deep-seated slide. ‐ A deep-seated slide may be triggered by an accumulation of sand, erosion (e.g. discharge of an outflow pipe or wave action) or some local small vibration in the earth crust. On the other hand, among the «positive» factors: ‐ The Quai des Chalands area appears to have no history of sliding, except between 1937 and 1946. ‐ The efficiency of the vibro-floated wall to stop a slope failure is not known. However observations during the vibro-floating operation indicated a very strong soil at the bottom of the vibro-floated area, where the cone penetrometer could not penetrate. With such layer beneath the vibro-floatation zone, the possibility of a slide going further inland should be low. ‐ Movements experienced during pile installation were not an indication of foundation instability, but were probably caused by the piling procedure used. ‐ The sheet pile driving near the Quai des Chalands did not trigger slides at the time of pile installation. An impact study was made to select the optimum solution for the continued operation of the oil terminal. The following consequences were considered: (1) loss of life, and loss of Pointe Odden to sea; (2) impact on environment due to damage of loading berth and oil leakage; (3) undermining of sand foundation at Quai des Chalands; (4) displacement of pile tops and anchors of the pier; (5) reduction of draught near the pier and Quai des Chalands; (6) impact on oil terminal activities. Remedial measures and their feasilibity were also considered: (1) on-site geotechnical reconnaissance and laboratory testing on soil samples; (2) careful dredging (difficult to achieve without triggering a slide, and not a permanent solution); (3) deep underwater compaction (e.g. vibrofloatation, chalk piling, grouting, chemical injection; (4) controlled blasting of the underwater slope.

Lessons learned

The slides seemed to occur every 15 to 20 years. They are a natural phenomenon due to the geology and geography of the area. For such natural hazard, given the impact analysis carried out and the uncertainty whether remedial measures would be partly or fully successful, continued surveillance of the coast and seabed, using the slope of the seabed as stability indicator, was deemed the optimum solution. The sliding on Cap Lopez was difficult to circumvent. In view of the sliding observations in the past, the fact that recent sliding occurred within the limits of the «worst case» scenario of the impact study, the positive and negative factors that could lead to sliding near the terminal installations, Elf made the engineering decision to continue surveillance of the coast and seabed, using the slope of the seabed as indicator of stability. This decision was helped by the fact that the risk of loss of human life was essentially nil. The overall risk to the oil terminal operations was considered to be tolerable with surveillance of the coast as the main tool to evaluate whether new remedial measures were needed. 8 8.1

THE ASHCROFT THOMPSON RIVER LANDSLIDES Description of the landslides

The town of Ashcroft is located on the east side of the Thompson River in southern British Columbia, northeast of Vancouver. The multiple landslide activity near Ashcroft has a very strong impact on freight transportation. (Bunce and Chadwick 2012; Bunce and Martin 2011; Bunce and Quinn 2012). Figure 26 presents an aerial view of part of the Ashcroft Thompson River and three recent landslides. Near the village of Ashcroft, more than 20 landslides have occurred, ranging in size from 10,000 m3 to 5 million m3. Figure 27 illustrates some of the reported landslide initiations and observed significant movements near the railway in the Ashcroft area. The movements can be slow and relatively small, but insidious, or they can be sudden, fast and very large. The stakeholders are the railway companies, Canadian Pacific and Canadian National, Transport Canada, the British Columbia Government (Environment and Transportation) and the Canadian Department of Fisheries and Oceans. Railways traverse valley slopes and can be exposed to numerous landslide hazards. Railways typically select one of three strategies to manage the risks associated with landslides: avoid the landslide, stabilize the landslide or implement monitoring and signal systems that indicate when the tracks may be unsafe (Bunce and Martin 2011).

Figure 25. Run-out distance vs slide volume for submarine slides (Edgers and Karlsrud 1982; additional data from Canals et al 2005))

Figure 26. Ashcroft Thompson River and three recent landslides (Bunce and Quinn 2012)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

cluded the magnitude and frequency of landslide activity and the rate of ground movement compared to the frequency of track maintenance. The impact of the failures was multi-faceted. In addition to the costs to the Canadian economy, the negative aspects included: potential for injury and death of locomotive operator and conductor, the impact on the environment, consequences of a derailment including the fate of the freight material, a prolonged service interruption resulting in a loss of Canada’s credibility as a reliable exporter, damage to key fisheries, impact to First Nations land claims, damage to adjacent land-use and irrigation for agriculture, flooding, damage upstream and downstream of the landslide. For the Ripley Landslide, since the track speed was 30 mph with no potential for a derailed locomotive to reach the river, the probability of a fatality was estimated as extremely low. The Ripley Landslide was known to be moving at a gradual rate that had had no influence on the safe operation of the railway for more than 60 years. The frequency of normal railway maintenance was sufficient to periodically realign the track such that the track speed could be maintained without compromising the safety of rail operations, despite periods requiring more frequent track maintenance. From an economic perspective the Ripley Landslide was costing the railways a minimal amount of maintenance and little or no reduction in operating efficiency. The primary successful landslide mitigation measure of the other landslide locations was the placement of an erosion-protection toe-berm of rip-rap into and along the river bank. However, although the cost of this method was attractive compared to other options the environmental, especially fisheries impact was considered significant. In the case of the Ripley Landslide, CP assessed its options and given that the effectiveness of stabilization was uncertain and costly, and the risk of catastrophic failure based on past performance of this landslide was low, a monitoring system was selected. The advantages of this concept were: the risk to train traffic was minimized; the cost was less than the least costly stabilization measures; the environmental impact was negligible in comparison to completing in-river works; and additional information about the behaviour of the landslide in response to external changes could be further investigated to identify means of stabilizing the landslide in the future if movement rates increase above tolerable levels. These advantages were offset by the disadvantage that although rail safety is ensured the reliability of the transportation system remains the same. In view of the uncertainties and the overwhelming extent of the potential consequences, CP invested in research and monitoring. The research investment included a Railway Ground Hazard Research Project (http://rghrp.com/), multi-year research grants and support for PhD and MSc studies on rock fall, landslides, climatic triggers, debris flows and risk analysis, a rail research laboratory (http://carrl.ca/) and strategic research partnership with universities, research organizations, and stakeholders. CP installed in 2008 a real time permanent Global Positioning System (GPS) on the Ripley Landslide located about 7.5 km south of Ashcroft to monitor ground movement and provide notification of significant track movement (Bunce and Chadwick 2012). The Ripley Landslide was known to have moved approximately 70 mm per year between 2008 and 2011. In view of (1) the high cost to stabilize 400,000 m3 of soil, (2) the environmental implication of attempting to stabilize the landslide without negatively changing the fishery in the Thompson River and (3) the uncertainty on the effectiveness of potential stabilizing measure; the decision was taken to monitor and respond rather than stabilize the landslide.

Figure 27. Reported landslide in Ashcroft area (Bunce and Quinn 2012).

8.2

Investigations

The soil consists of disturbed glacio-lacustrine clay and silt, and the failure seemed to follow complex mechanisms with irregular wedge formation. The geological and hydrogeological settings were also complex, with alluvial fans and fractured bedrock (Bunce and Quinn 2012). The geotechnical investigation failed to identify a trigger for increased movement. Given the long period of gradual movement it appeared that the slope was in a alternating cycle of being unstable and stable due to erosion and or groundwater conditions and small increments of movement. As part of the planning of mitigation work and the management of the landslide activity and operative safety of the railroad, knowledge gaps were identified:  Subsurface conditions outside and between landslides.  Stress-strain behaviour of the materials involved in failure.  Realistic model for new or reactivated landslides.  Contribution of river drawdown, erosion and infiltration.  Erosion by the river.  Effect of weather and climate, and changes thereof.  Effects of topography.  What are the tolerable movement limits?  Local water balance. 8.3

Analyses of the slides

Some Ashcroft Thompson River landslides are known to have moved at rates of several meters per day including the North Landslide in 1881 (Stanton 1898) and the Goddard Landslide in 1982 (Fig.91). The Ashcroft Sub, Mile 50.9 Landslide and the active portions of the North Landslide and the South Landslide are known to be currently moving at rates of 10 to 30 mm year. The causes of the landslides were multiple, and at times difficult to assess, which make the prediction of an oncoming landslide as railroad traffic is planned very uncertain. The causal factors include (Bunce and Chadwick 2012):  weak glacio-lacustrine silt and clay;  incision of the Thompson River;  upward seepage pressures;  low strength on pre-existing failure surfaces;  river level appears to exert a controlling influence;  infiltration from irrigation. There was also relative little information on the success and/or failure of past remedial measures. 8.4

Risk management

Bunce and Martin (2011) developed a procedure to manage the railroad risk associated with landslides. Factors considered in-

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The monitoring system had three GPS antennae on the landslide and one stationary reference antenna. Accuracy for the longer term 24 hour averaged data was better than 5 mm. The landslide monitoring data revealed that the landslide moved fastest in the spring prior to the highest river levels (Bunce and Chadwick 2012). The GPS landslide monitoring system could provide real time warning to approaching trains of ground movement and possible track misalignment. Figure 28 illustrates on a semi-log scale the costs associated with landslides in the Ashcroft Thompson River area as a function of the return period of the landslides.

9.1

Description of the slide

During the night of 26th March 2008, a rock slope failed and destroyed a new apartment complex in Aalesund, Norway. About 1400 m3 of rock rammed in the building. The lower floors were completely collapsed and set on fire. The entire building was displaced by several meters. There were 20 persons in the building at the time of the collapse, and 5 persons, all from the lower floors, died. The other residents were not injured. The accident was tragic and affected many in Norway because of its actuality and because it hit “close to home”, in the privacy of one’s apartment. Figure 29 illustrates the sliding of the rock mass in the building already on fire and attended by firemen. The cause of the rock slide was the presence of a plane of weakness filled with clayey material, and the creation of additional fissures by the blasting during the preparation of the site before the construction started. Figure 30 illustrates the plane of weakness. 9.2

Lessons learned

The accident could have been avoided if a proper site investigation had been carried out. In particular, geophysical methods should have been used both before and after the blasting for the site preparation. Before the blasting, the weakness plane would have been discovered, and bolting would then have been undertaken. The geotechnical/engineering geology site investigation report was insufficient. This omission cost the lives of five persons. After the accident, engineering geologists studied the stability of the rock for all neighbouring buildings, and the rock wall has been stabilized.

Figure 28. Cost of landslides for the Ashcroft Thompson River railroad (Bunce and Quinn 2012).

The total cost was calculated as the sum of the cost of railway service interruption and the cost of the of railway revenue. Figure 26 indicates that as soon as railway service interruption exceeded about two to three days, the total cost became exponential. The curves show that for a landslide with return period of only 20 years, the total cost reach an astronomical sum of 800 MCAD or more. Preventing the smaller, more frequent landslides became therefore a priority. On the basis of the diagram in Figure 27, it would seem justified for the stakeholder to spend about 5 to 10 MCAD in mitigation measures to avoid the damage due to a 10-year return period landslide. This was translated into a recommendation to continue research with an additional 0.5 to $1.5 MCAD/year and to do the stabilization of known landslides at a cost of 2 to 5 MCAD per landslide per year. 8.5

THE AALESUND SLIDE

Lessons learned

This case study presents an excellent example of risk management and decision-making under uncertainty, where the potential negative consequences on both short and longer term were considered and a compromise solution was selected. In one case, the decision was made to reduce the risk with monitoring and warning to avoid the cost of mitigation with uncertain outcome, and to avoid the environmental impact of stabilizing the landslides. A GPS landslide monitoring system was used to notify approaching trains if the ground movement has exceeded a threshold that rendered the track impassable. When the implications of avoiding or stabilizing landslides are significant, this can be a viable risk reduction strategy. However, this approach did not reduce the likelihood of a prolonged service interruption, with the ensuing costs. With a view towards future improvements, the stakeholders decided to invest in research and to quantify how much money they could spend on mitigation compared to the cost on of letting the landslides occur, in order to document the costeffectiveness of mitigation and monitoring.

Figure 29. The Aalesund rock slide

Figure 30. Sliding of rock mass on weakness plane (NGI files)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

10 THE STOREGGA SLIDE

required for the exploitation of the field; and (2) Can smaller slides be triggered on the steep slopes created by the Storegga slide, and if so, would they endanger the planned offshore installations to recover the gas resources.

10.1 Description of the slide The Storegga slide in the Norwegian Sea is one of the largest known submarine slides on earth. The head wall of the slide scar is 300 km long. About 3500 km3 failed from the shelf edge, sliding out as far as 800 km in water depths as deep as 3000 m (Fig. 31). The failure started probably some 200 km downhill and crept rapidly up slope as the headwalls failed and slipped down towards the deep ocean floor. At the same time, the mass movement generated a huge tsunami that reached the shores of, among others, Norway, Scotland and the Shetland Islands. The sizable gas resources at the Ormen Lange filed are located in the scar left by the giant underwater slide, beneath a relatively chaotic terrain created by the slide 8,200 years ago. The Storegga slide was the subject of a large integrated study for the safe development of the deepwater gas field on the North Atlantic continental margin. In addition, the SEABED project was launched by the partners of the Ormen Lange field (Norsk Hydro ASA, A/S Norske Shell, Petoro AS, Statoil ASA, BP Norge AS and Esso Exploration and Production Norway AS) with the aim of improving the knowledge of the seafloor morphology, the shallow geology, and the potential hazards and risks associated with the area. The project is an excellent example of the interweaving of research and practice and the cooperation of academia and industry.

10.2 Soil parameters The reader is referred to Solheim et al (2005a; b); Kvalstad et al (2005 a;b); Kvalstad (2007); Nadim et al (2005b) and the special issue of Marine and Petroleum Geology (Volume 22, No 1 and 2) for an account of the slide and a summary of the studies by the parties involved. 10.3 Analysis of the landslide Based on the studies in the SEABED project, the triggering and sliding mechanics used the observed morphology and the geotechnical characteristics of the sediments. The average slope angle of the seafloor was only 0.6 to 0.7°. The geotechnical properties indicated shear strengths far above those required to explain a failure. However, the geophysical observations, especially seismic reflections profiles in the upper parts of the slide scar, provided strong indications that the failure developed retrogressively (Fig. 32). Using the retrogressive slide model as working hypothesis, several scenarios of sources of excess pore pressures were considered, including (1) earthquake-induced shear strain generating excess pore pressures, (2) melting of gas hydrates releasing methane gas and water, (3) shear straininduced contraction with pore pressure generation and strainsoftening, and (4) rapid deposition. The studies concluded that the most likely trigger was an earthquake destabilizing a locally steep slope in the lower part of the present slide scar. The retrogressive process continued up-slope until conditions improved with stronger layers associated with the consolidation of the shelf sediments during glacial times. Once the instability started, excess pore pressures already generated during rapid sedimentation under the last glaciation were an important contribution to the large slope failure (Bryn et al 2005).

Figure 32. Bathymetry and seismic profiles in the upper headwall at Ormen Lange (Kvalstad et al 2005a).

Excess pore pressures still exist at the site, as demonstrated by in situ monitoring (Strout and Tjelta 2005). The excess pore pressures recorded in several locations and at several stratigraphic levels support the depositional role in the Storegga failure proposed by Bryn et al 2005. The seismic studies by Bungum et al 2005 showed that strong, isostatically induced earthquakes had occurred earlier along the mapped faults at the site. Stress transfer induced earthquakes had also probably taken place earlier. Bungum et al also suggested that multiple strong earthquakes with extended duration most likely occurred and could be the potential trigger for the Storegga slope instability. The tsunami generating potential of submarine slides is today widely recognized. The tsunami studies indicated that the field observations of run-up fitted will the retrogressive slide model with a velocity of 25-30 m/s, and time lags of 15-20 s between individual slide blocks (Bondevik et al 2005). The slide mass involved in the tsunami generation modelwas 2,400 km3.

Figure 31. The Storegga slide, 8,200 years BP.

The design questions that needed to be answered were: (1) Can a new large slide, capable of generating a tsunami, occur again, either due to natural processes or through the activities

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behavior and potential hazards and risks associated with the area. The interweaving of research and practice, the cooperation of academia and industry and the integration of the geodisciplines were essential for gaining an understanding of the past slide and providing the possibility to develop the gas field.

Figure 33 presents an illustration to explain the sedimentation process leading to failure, which supports the hypothesis that major slides have occurred in the Storegga area on a semiregular basis, related to the glacial/interglacial cyclicity. The bottom illustration in Figure 33 (denoted 1) gives the last interglacial with deposition of soft marine clays. The middle illustration (denoted 2) presents the last glacial maximum (LGM) with the ice at the shelf edge and deposition of glacial sediments. The top illustration (denoted 3) presents the topography after the Storegga slide. Dating (BP, before present) is given for each illustration. The illustration denoted 3 also shows two older slide scars that were filled with marine clays. The slip planes were found in seismically stratified units of hemipelagic deposits and the thick infill of stratified sediments indicate a late glacial to early interglacial occurrence of slides (Bryn et al 2005). The soft fine-grained hemipelagic deposits were rapidly loaded by coarser glacial deposits during the short glaciations period. Excess pore pressures were a destabilizing factor. The hypothesis of strong earthquake shaking was retained to start the underwater slide. After the earthquake initiated the movement, the slide continued retrogressively by back-stepping up the slope where the pore pressures were already high. The mass movement was further facilitated by the release of support at the toe. The stability of the present situation at Ormen Lange was evaluated by Kvalstad et al 2005b. The conclusion was that an extremely strong earthquake would be the only realistic triggering mechanism for new submarine slides in the area. The annual probability of third party damage was also investigated and found to be extremely low (Nadim et al 2005b). The project team therefore concluded that developing the Ormen Lange gas field could be done safely.

11 LANDSLIDE RISK MANAGEMENT 11.1 Landslide prevention in Drammen In Norway, the hazard ere estimated on the basis of simple theoretical evaluations of the potential area that can be involved in a quick clay slide, in combination with back-calculations of a number of historical quick clay slides (Aas 1979). The assessment of the risk associated with slides in sensitive clays in Norway is a semi-quantitative approach developed for the Norwegian Water Resources and Energy Directorate (NVE). Slide areas are classified according to “engineering scores” based on an evaluation of the topography, geology and local conditions (to qualify hazard) and an evaluation of the elements at risk, persons, properties and infrastructure exposed (to qualify consequence). The risk score to classify the mapped areas into risk zones is obtained from the relationship RS = HWS  CWS, where RS is the risk score, HWS is the weighted hazard score and CWS is the weighted consequence score. The risk matrix is divided in five risk classes. Guidelines for the implementation of the risk matrix are administered by NVE. In practice, the approach is used to make decisions on required mitigation measures to reduce the risk. The approach is simple and makes room for engineering experience and judgment. For detailed regional planning, slope stability calculations need to be made. The approach has been described in detail in Gregersen 2005; Lacasse et al 2003; Lacasse and Nadim 2008; and Kalsnes et al 2013. A similar procedure has been developed for sensitive clays in Québec (Thibault et al 2008), reflecting the experience with large retrogressive slides in Québec. An example of the management of risk based on the above scores is the preventive actions set in place in Drammen. The city of Drammen, along the Drammensfjord and the Drammen River, is built on soft sensitive clay. Stability analyses were done in an area close to the centre of the city, and indicated that some areas did not have satisfactory safety against a slope failure. Based on the results of the stability analyses and the factors of safety (FS) obtained, the area under study was divided into three zones, as illustrated in Figure 34: – Zone I FS satisfactory – Zone II FS shall not be reduced – Zone III FS too low, area must be stabilised Figure 35 illustrates the mitigation done in Zone III: a counter fill was immediately placed in the river to support the river bank, and the factor of safety checked again. The counter fill provided adequate stability. In Zone II, no immediate action was taken, but a ban was placed on any new structural and foundation work without first ensuring increased stability. Figure 36 illustrates required actions:  if an excavation is planned, the clay will have to be stabilised with e.g. anchored sheetpiling or soil stabilisation, for example lime-cement piles;  if new construction is planned, the engineer needs to check the effects of the change on the stability down slope: e.g. adding a floor to a dwelling may cause failure because of added driving forces; or new piling up slope can cause an increase in pore pressures and a driving force on the soil down slope. With such an approach, focus is set on the need for mitigation rather than as the risk and potential for failure.

Figure 33. Deposition and sliding processes (Bryn et al 2005).

10.4 Lessons learned The documentation of the feasibility of pipeline installation across the Storegga slide would not have been possible without the integrated inter- and cross-disciplinary study of the development, now without the conscious effort to improve knowledge on seafloor morphology, shallow geology, geotechnical

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

11.2 The SafeLand Project

Figure 34. Classification of hazard zones in Drammen (FS = safety factor) (Gregersen, 2008).

Figure 35. Mitigation in hazard Zone III in Drammen.

The need to protect people and property in view of the changing pattern of landslide hazard and risk caused by climate change and changes in demography, and the need for societies in Europe to live with the risk associated with natural hazards, formed the bases for the 2009-2012 SafeLand project “Living with landslide risk in Europe: Assessment, effects of global change, and risk management strategies”. SafeLand was an integrating research project under the European Commission’s 7th Framework Programme. The project involved 27 partners from 12 European countries, and had international collaborators and advisers from China, India, USA, Japan and Hong Kong. SafeLand also involved 25 EndUsers from 11 countries. SafeLand was coordinated by NGI’s Centre of Excellence “International Centre for Geohazards (ICG)” (http://safeland-fp7.eu/). Nadim and Kalsnes (2014) present the results of the project in more detail. The objectives achieved in the SafeLand project include: ‐ Guidelines related to landslide triggering processes and runout modelling. ‐ Development and testing of empirical methods for predicting the characteristics of threshold rainfall events for triggering of precipitation-induced landslides, and development of an empirical model for assessing the changes in landslide frequency (hazard) as a function of changes in the demography and population density. ‐ Guidelines for landslide susceptibility, hazard and risk assessment and zoning. ‐ New methodologies for the assessment of physical and societal vulnerability. ‐ Identification of landslide hazard and risk hotspots in Europe. The maps show the location of the areas with highest landslide risk and allow a ranking of the countries by exposed area and population. ‐ Simulation of regional and local climate change over regions of Europe at spatial resolutions of 10 x 10 km and 2.8 x 2.8 km. The simulations were used for an extreme value analysis of trends in heavy precipitation events, and subsequent effects on landslide hazard and risk. ‐ Guidelines for the use of remote sensing, monitoring and early warning systems. ‐ Development of a prototype web-based "toolbox” of innovative mitigation measures. The toolbox does a preliminary assessment of the appropriateness of the measures and a ranking of over 60 structural and non-structural landslide risk mitigation options. ‐ Case histories and "hotspots" of European landslides were collected and documented. Data for close to fifty potential case study sites (Italy, France, Norway, Switzerland, Austria, Andorra, and Romania) were compiled. Almost all types of landslide and types of movement were represented. ‐ Stakeholder workshops and participatory processes to involve the population exposed to landslide risk in the selection process for the most appropriate risk mitigation measure(s). 11.3 Slope safety in Hong Kong

Figure 36. Mitigation and preventive measures in Zone II in Drammen.

The best example worldwide of a comprehensive and effective program of risk management for landslides is probably the slope safety program administered by the Geotechnical Engineering Office (GEO) of the Civil Engineering and Development Department in Hong Kong. The Slope Safety System has seven main targets: ‐ Improve slope safety standards, technology, and administrative and regulatory frameworks. ‐ Ensure safety standards of new slopes.

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‐ ‐ ‐ ‐

formation within each of the organizations was insufficient; control of the information given to the media was problematic (it was not possible to check the wording with the journalists); making notes and looking at maps in a high pressure context proved very difficult; how could the private actors doing emergency work/repair ensure that they had a contract (and would be paid for their work); the authorities should create an emergency group, and invite wide participation; a logistical and communication tool should be developed for crisis situations and made available to all stakeholders; the need for frequent preparedness exercise, as the people change in each of the organisations involved; and some of the routines in the governmental handbooks needed to be updated.

Rectify substandard Government man-made slopes. Maintain all Government man-made slopes. Ensure that owners take responsibility for slope safety. Promote public awareness and response in slope safety through public education, publicity, information services and public warnings. ‐ Enhance the appearance and aesthetics of slopes Hong Kong has a history of tragic landslides. Since 1947, more than 470 people died, mostly as a result of failures associated with man-made cut slopes, fill slopes and retaining walls. Today, the risk to the community has been greatly reduced by concerted Government action. On average, about 300 incidents affecting man-made slopes, walls and natural hillsides are reported to the Government each year. To reduce landslide risk, GEO assures the operation of a 24hour year-round emergency service by geotechnical engineers to protect the public, investigates all serious landslides, and with this experience continuously improves its knowledge and standards. GEO also audits the design and supervision of construction of all new slopes to ensure that they meet the required safety standards, upgrades “substandard” slopes based on a priority matrix and takes steps to ensure that private owners take responsibility for their own slopes through safety screening. Natural terrain studies and risk mitigation actions are carried our continuously. In addition, GEO undertakes extensive public education on personal safety precautions in order that the community can be better informed on how to protect themselves during periods of intense rainfall when landslides are likely to occur. There is also a program to assess squatter villages for clearance of squatter huts and to provide guidance to the residents on landslide risk and self-protection. Complementary to enhancing the stability of slopes, GEO also gives priority to beautifying the slopes, either by making them look as natural as possible or blending them with the surroundings. Technical guidelines have been issued on good practice in landscape treatment and bioengineering for slope work.

11.5 Recent developments 11.5.1 Mapping tool for quick clays Geophysical methods, especially Electric Resistivity Tomography, have emerged as reliable tools for quick clay mapping, as witnessed at several quick clay sites in Norway (Pfaffhuber et al 2012). When combined with borehole data and electric resistivity cone penetrometer data, the methods are also cost-effective. Geochemical analysis also demonstrated that changes in resistivity are directly related to changes in clay salt content, and related to clay sensitivity. The usefulness of geophysical investigations offshore for the determination of the soil characteristics and correlation of layers has already been demonstrated in offshore work in deep waters. 11.5.2 Mitigation The United Nations’ International Decade for Natural Disaster Reduction (1990-2000) to reduce loss of life, property damage and social and economic disruption caused by natural disasters, was the start of international concerted actions. Mitigation and prevention of the risk posed by landslides, however, did not attract widespread and effective public support in the past. The situation has changed dramatically during the past decade, and it is now generally accepted that a proactive approach to risk management is required to significantly reduce the loss of lives and material damage associated with natural hazards. The wide media attention on major natural disasters during the last decade has clearly changed people's mind in terms of acknowledging risk management as an alternative to emergency management. A milestone in recognition of the need for natural disaster risk reduction was the approval of the "Hyogo Framework for Action 2005-2015: Building the Resilience of Nations and Communities to Disasters" (ISDR 2005). This document, approved by 164 UN countries during the World Conference on Disaster Reduction in Kobe in January 2005, defines international working modes, responsibilities In the 20th century, the economic losses from natural hazards were greatly underestimated, the awareness of hazards and risk was insufficient, and the mitigation and regulation to avoid damage and loss was inadequate. Since 2005, the awareness of the need for mitigation of natural hazards has greatly increased. On the other hand, since the 80's, hazard and risk assessment of the geo-component of a system has gained increased attention. The offshore, hydropower, nuclear and mining industry were the pioneers in applying the tools of statistics, probability and risk assessment. Gradually, environmental concerns and natural hazards started implementing hazard and vulnerability assessment. Nowadays the notion of hazard and risk is a natural question in most geotechnical engineering aspects and even project management.

11.4 Preparedness NVE organized in 2010 an exercise in landslide preparedness, and the Norwegian Directorate for Civil Protection and Emergency Planning (DSB) in 2013. The first exercise simulated a quick clay landslide of national dimension with fatalities. The second assembled a group of experts to establish the premises for the national risk that could be posed by quick clay slides. Worst case scenario, estimates of hazard and vulnerability and valuation of the consequences were discussed by the different stakeholders involved. The results will become available in June 2013. The NVE simulation in 2010 was made as realistic as possible with the participants not knowing beforehand what to expect and having one party simulate fatalities. The participants were briefed of the exercise ahead of time, but they did not know the details of what was to happen. A majority of stakeholders were invited, including authorities, police, private actors and media, in addition to the technical instances required in such emergency situation. The exercise aimed at improving the parties’ ability to respond under pressure in a complex context, and making decisions under critical conditions. The emergency routines, information channels and response tools in each of the participating organizations were tested. The exercise also tested who took responsibility for the decisions made, and whether the parties had the same understanding of the respective responsibility and roles. An evaluation report was prepared with, among others, lessons learned: the respective roles and responsibility should be more clearly defined and communicated to all parties; not everyone received the required information in time and internal in-

12 THE SIGNIFICANCE OF SAFETY FACTOR The factor of safety against instability is a measure of how far a slope may be from failure. Factors of safety are applied to compensate for uncertainties in the load, resistance and parameters

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Increase in driving forces  increase in external loads;  pore pressure in weakness zones;  frost in fissures;  increase in unit weight due to rainfall;  excavation or erosion at toe of slope;  lowering of water table;  earthquake loading. Figure 39 illustrates the development of a progressive failure in a strain-softening soil. The displacement along the slip surface varies between the toe and the top of the slope. One therefore needs to include strain compatibility in the analyses.

thereof, and model used for the calculation. The factor of safety is often expressed as the ratio of the resisting forces to the driving forces. For a slope to be stable, the stabilizing forces (moments) should be larger than the driving (destabilizing) forces. If there was no uncertainty in the safety factor, a safety factor of 1.05 would be sufficient. However there is uncertainty in nearly all the parameters that enter the analysis of the stability of a slope. There will therefore always be a finite probability that the slope will fail. Defining the level of the finite probability that is tolerable is the challenge. The geotechnical engineer should provide insight in this discussion. Figure 37 illustrates with probability density functions the notion that factor of safety alone is not a sufficient measure of the margin of safety. The figure gives the probability density function (PDF) for two slopes. The first has a central FS (or SF based on mean values) of 1.4 and a probability of failure, Pf, of 10-4 per year. The probability of failure is illustrated by the area ace where the factor of safety can be less than unity. The second slope has a more diffuse PDF, has a higher FS of 1.8, but also a higher probability of failure, Pf, of 10-3 per year (zone aed). In order to select a suitable factor of safety, one therefore needs to estimate the uncertainties involved. There exists no relationship between safety factor based on limit equilibrium analysis and annual probability of failure. Any relationship would be sitespecific and depend on the uncertainties in the analysis.

Figure 38. Brittle and strain-softening material

Figure 37. Factor of safety and probability of failure.

The safety factor should not be a constant deterministic value, but should be adjusted according to the level of uncertainty. Ideally, given time and money, one could calibrate the required safety factor for different classes of slopes, soils and failure types that would ensure a target annual probability of failure of for example 10-3 or 10-4 per year. In most cases, after a slide has occurred, it is difficult to determine a unique trigger for the slide. Whether a material is ductile or brittle (Fig. 38) is a very important factor that causes uncertainty in the shear strength to use and how well the failure mechanism of the slope is captured and modelled in the stability analyses. The strain-softening of the brittle material is especially problematic, as it will show reduced resistance once a threshold shear deformation has been exceeded (b in Fig. 38). The stability conditions are especially difficult to analyse, e.g. for the Vestfossen and the Kattmarka landslides. Landslides can be triggered by natural causes (geological, geomorphological or hydrological/meteorological processes) or by human intervention. Triggers can be: Reduction of soil resistance:  increase in pore pressure (artesian pressure, rainfall etc);  cracks on top of slope;  swelling;  chemical changes;  reduction of shear strength towards residual strength;  creep deformations;  vibrations with temporary increase in pore pressure;  rainfall (intensity and duration).

Figure 39. Illustration of progressive failure.

For sensitive clays in Norway, the average mobilized shear stress along different slip surfaces is compared with the corresponding characteristic average shear strength divided by a material coefficient, γM. The requirement in Norway for the material coefficient γM is 1.4 (NVE 2011), and this γM should be used in design. However, for an existing (standing) slope with a material coefficient less than 1.4, NVE requires that the back-calculated material coefficient should be increased, but not necessarily to as much as 1.4. Figure 40 illustrates the NVE requirement. The required γM is a function of the initial back-calculated material coefficient and the improvement required. The improvement required depends on the hazard class (NVE 2011). The upper diagram in Figure 40 gives the minimum required increase in material coefficient γM, (in %), for two levels of slope improvement: “Substantial improvement required” and “Improvement required”. The lower diagram provides the resulting

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required γM. The improvements are to be made through topographical modifications. As an example (lower diagram), for an initial γM of 1.2, an improvement from γM = 1.2 to γM = 1.26 and 1.29 is required by NVE (2011) for the two levels of improvement specified. Standing slopes with a material factor γM of 1.0 require an improvement up to γM = 1.10 and 1.15 for the two levels of “improvement” specified by NVE. The reason for allowing a material coefficient less than 1.4 is that the fact that the slope is standing today is a confirmation that the slope has a material coefficient of at least 1.0. Any improvement therefore represents a real gain to the present safety of the slope. The NVE requirement needs to be satisfied for all potential slip surfaces. For sensitive clays, the peak undrained shear strength is reduced in limit equilibrium analyses to account for strainsoftening at large shear strains. A reduction of 10 to 15% in the peak shear strength in triaxial compression, triaxial extension and direct simple shear may be adequate, as discussed under the Vestfossen case study. However a reduction factor should probably be developed for different categories of clays and slip surfaces..

The geotechnical engineer should be aware that it is more correct and safer to ensure that slopes have the same probability of failure rather than the same factor of safety. Mitchell and Kavazanjian (2007) presented “Geo-engineering Engineering for the 21st Century”. On request from the National Science Foundation in the USA, an expert committee suggested a vision for how geo-engineering could continue to address societal needs in the 21st century, and identified emerging technologies that could contribute to this vision. Mitigation of natural hazards was one of the areas identified. Emerging technologies included:  An improved ability to “see into the earth” and interpret geophysical surveys.  Improved sensing and monitoring, more reliable instrumentation, enhanced data acquisition, processing and storage, and appropriate information systems.  Improved ability to characterize the spatial variability of soil properties and the uncertainty in the assessments made. In addition, inter- and cross-disciplinary problem-solving is essential for advancing in the practice of geo-engineering. More emphasis must be placed on inter-disciplinary collaboration, in research, consulting and education. The expertise of geotechnical engineers is essential for meeting the challenge of protecting society, worldwide. Safety and life quality depends on our profession. We must however avoid being unaware of the impact of the work we do as engineers. To paraphrase Siegel (2010): civil engineers built the countries we live in. Civil engineers make a difference in the world: “When we flip a switch, the lights come on. When we turn on the tap, we trust that the water is clean and potable. When we drive home from work, we trust the roads will not collapse”. Over the last 100 years, life expectancy has doubled. The main factor has not been advances in medicine, but advances in clean water technology and sanitation. Civil engineers are solving the world’s problems every day. In closing this 8th Terzaghi Oration, I return to Professor Ralph B. Peck, who early in his career, already defined the civil engineer’s role in a most adequate manner. The key to success and happiness, in his view, was “[...] a love of civil engineering, which, at its core, seeks to do 'good works' for humanity”. In view of today’s needs and our profession’s evolution, Ralph could not have been more right.

1.5 Substantial improvement required

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14 ACKNOWLEDGMENTS

1.3

The author wishes to thank President Jean-Louis Briaud for selecting her to be the 2013 Terzaghi Orator. The author is also thankful to NGI for the opportunities it gave her throughout her career. The author is indebted to many colleagues who provided data and information for the case studies, especially Dr Hans-Peter Jostad, Håkon Heyerdahl, Bjørn Kalsnes Dr Maarten Vanneste, Arnstein Aarset, Dr Farrokh Nadim, Odd Gregersen, Dr Andi A Pfaffhuber,Tim Gregory and Dr Kaare Höeg, all from NGI. The assistance and prompt reply to my questions from Professor Steinar Nordal, from NTNU, Dr Denis Demers and his colleagues at Ministère des Transport du Québec and Dr Chris Bunce from Canadian Pacific are also greatly appreciated.

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Figure 40. Required increase in material coefficient (top diagram, NVE 2011) and resulting required material coefficient (lower diagram) for an existing (standing) slope.

15 REFERENCES Aas G. 1979. Skredfare og arealplanlegging. Vurdering av faregrad og sikringstiltak. Ullensvang Hotell, Lofthus i Hardanger. Norske sivilingeniørers forening. Oslo: NIF, 1979. 1. b. Berre T., Lunne T., Andersen, K.H., Strandvik, S.O., Sjursen 2007. Potential improvements of design parameters by taking block samples of soft marine Norwegian clays. Canadian Geotechnical Journal.44 698-716. Bondevik S., Løvholt F., Harbitz C., Mangerud,J., Dawson A and Svendsen J.I. 2005. The Storegga Slide tsunami-comparing field observations with numerical simulations. Marine and Petroleum Geology 22, 195-208.

13 CONCLUSIONS The geotechnical engineer’s role is not only to act as technologist providing judgment on factors of safety. The role has evolved to providing input in the evaluation of hazard, vulnerability and risk associated with landslides. The geotechnical profession should be increasingly perceived as reducing risk and protecting people.

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013 Lacasse S. 2013. Wizard for GeoSuite software. Paper to 2013 Canadian Geotechnical Conference. Montréal. October 2013. Locat A. 2007. Etude d’un étalement latéral dans les argiles de l’est du Canada et de la rupture progressive – Le cas de Saint-Barnabé-Nord. MSc Thesis, Université Laval, Québec. 262 p. Locat A., Leroueil S., Bernander S., Demers D., Locat J. and Ouehb L. 2008. Study of a lateral spread failure in an Eastern Canada clay deposit in relation with progressive failure. The Saint-Barnabé-Nord slide. Geohazards IV, Québec 89-96. Locat P., Fournier T., Robitaille D. and Locat A. 2011. Glissement de terrain du 10 mai 2010, Saint-Jude, Montérégie – Rapport sur les caractéristiques et les causes. Ministère des Transports du Québec, Service de la géotechnique et de la géologie, Rpt MT11-01, 101 p. Locat P., Demers D., Robitaille D., Fournier T., Noël F., Leroueil S., Locat A. and Lefebvre G. 2012. The Saint-Jude landslide of May 10, 2010, Québec, Canada. 11th International and 2nd North American Symposium on Landslides, Taylor & Francis London. 635-640. Marine and Petroleum Geology (2005). Thematic Set Ormen Lange. A. Solheim, P. Bryn, k. Berg, H.P. Sejrup and J. Mienert (eds). V 22; 12. Jan.-Feb.2005. 318p. Mitchell J.K. and Kavazanjian E. 2007. Geoengineering Engineering for the 21st Century. Geo-Strata July/Aug 2007, 14-18. Munich RE (2011). www.munichre.com/app_pages/touch/ naturalhazards/@res/pdf/NatCatNews/2011_01_03_munich_re_NatCatSERVI CE_en.pdf?1. Nadim F., Kvalstad T.J. and Guttormsen T.R. 2005. Quantification of risks associated with seabed instability at Ormen Lange, Marine and Petroleum Geology, 22: 311-318. Nadim F. and Kalsnes B.G. 2014. Progress of Living with landslide risk in Europe. Plenary Lecture. World Landslide Forum 3 Landslide Risk Mitigation: Towards a Safer Geo-Environment. Beijing, China. NGI 2012. Effekt av progressive bruddutvikling for utbygging i områder med kvikkleire. A2 Tilbakeregning av skred. NGI Report 20092128-00-5-R. 1st June 2012. Nordal, S., Alén C., Emdal A., Jendeby L., Lyche E. and Madshus.C 2009. Skredet i Kattmarkvegen i Namsos 13. mars 2009. Rapport fra undersøkelsesgruppe satt ned av Samferdselsdepartementet. NVE (2011): Retningslinjer 2/2011. Flaum- og skredfare i arealplanar. Rev.15.ISSN:1501-9810. Pfaffhuber A.A, Bazin S. and Helle T.E. 2013. An integrated approach to quick-clay mapping based on resistivity measurements and geotechnical investigations. 1st Intern. Wksp Landslides in Sensitive Clays. Québec. October 2013. Siegel B. 2010. We must not sell ourselves short: Engineering is an Honorable Profession. Geo-Strata July/Aug 2010, 44-47. Solheim, A, K. Berg, C.F. Forsberg and P. Bryn (2005a). The Storegga Slide complex: repetitive large scale sliding with similar cause and development. Marine and Petroleum Geology, 22: 97-107. Solheim, A., P. Bryn, H.P. Sejrup, J. Mienert and K. Berg (2005b). Ormen Lange – An integrated study for safe development of a deepwater gas field within the Storegga Slide Complex, NE Atlantic continental margin: Executive summary, Marine and Petroleum Geology, 22: 1-9. Strout, J.M. and T.I. Tjelta (2005). In situ pore pressures: what is their significance and how can they be reliably measures? Marine and Petroleum Geology, 22: 275-286. Thibault C., Potvin J. and Demers D. 2008. Development of a quantitativee approach for evaluating and managing the risk associated with large retrogressive slides. GeoEdmonton – Canadian Geotechnical Conference. September 2008.et al 2008 UNDP (United Nations Development Programme) (2004). "Reducing Disaster Risk – A Challenge for Development." Bureau for Crisis Prevention and Recovery, New York, 146 pp. UNISDR (2009). "Global assessment report on disaster risk reduction (GAR 2009)." ISBN/ISSN: 9789211320282, www.preventionweb. net/english/hyogo/gar/ Varnes D.J. (1978) Slope Movement Types ans Processes. In: Landslides: Analysis and Control (R.L. Schuster ans R.J. Krizek, ed.), Special Report 176, TRB, National Research Council, Washington, D.C., pp.11-33.

Bryn P., Berg K., Solheim A., Kvalstad T.J. and Forsberg C.F. 2005. Explaining the Storegga Slide. Marine and Petroleum Geology 22, 11-19. Bunce C.M. 2008. Risk estimation for railways exposed to landslides. PhD thesis. University of Alberta. Edmonton, Canada. 450 p. Bunce C.M. and Chadwick I. 2012. GPS monitoring of a landslide for railways. 11th Intern. Symposium on Landslides (ISL). Banff. AL. Bunce C.M. and Martin C.D. 2011. Risk estimation for railways exposed to landslides. 5th Canadian Conference on Geotechnique and Natural Hazards. Kelowna, BC Bunce C.M. and Quinn P. 2012. Ashcroft Thompson River Landslides Impact on Freight Transportation. Canadian Risk and Hazard Network Conf. http://www.crhnet.ca/pastsymposiums/2012/2012.htm Bungum, H., Lindholm C. and Faleide J.I. 2005. Postglacial seismicity offshore mid-Norway with emphasis on spatio-temporal-magnitudal vatiations. Marine and Petroleum Geology 22, 137-148. Canals M. et al 2004. Slope dynamics and impact form seafloor and shallow sub-seafloor geophysical data: case studies from the COSTA project. Marine Geology.213 9-72. Dibiagio E. 2013. Field instrumentation–the link between theory and practice in geotechnical engineering. 7th Intern. Conf. Case Histories in Geotechnical Engineering. May 1-4 2013. Chicago. No RBP-6. Edgers L. and Karlsrud K. 1982. Soil flows generated by submarine slides-case studies and consequences. Proc. 3rd Intern. Conf. Behaviour of Off-Shore Structures. Cambridge Mass. Vol. II. pp. 425-437. Fornes P, Jostad H.P. 2013. A probabilistic study of an inclined slope in sensitive clay using FEA. Subm. ComGeoIII 3rd Intern. Symp. Computational Geomechanics. Krakow. 21-23 Aug. Gregersen O 2008. Kartlegging av skredfarlige kvikkleireområder. NGM 2008. 15th Nordic Geotechnical Conf. Sandefjord. 178-186. Grimstad G., Andresen L., Jostad H.P. (2011). NGI-ADP: Anisotropic shear strength model for clay. International Journal for Numerical and Analytical Methods in Geomechanics, 36, 4, pp. 483-497. Grimstad G, Jostad H.P. (2011a). Stability analyses of quick clay using FEM and an anisotropic strength. NGM 2012. 16th Nordic Geotechnical Conf. Copenhagen. 2:675-680. Grimstad G. and Jostad H.P. (2011b). Effect of progressive failure in sensitive clays. Fjellsprengningsteknikk - bergmekanikk geoteknikk. Oslo 2011. Foredrag 38. 12p. Grimstad G. and Jostad H.P. (2012). Stability analyses of quick clay using FEM and an anisotropic. Published in NGM 2012, 16th Nordic Geotechnical Meeting, Copenhagen, 2, pp 675-680. Grimstad G., Jostad H.P. and Andresen L. (2010). Undrained capacity analyses sensitive clays using the non-local strain approach. Proc. 9th HSTAM International Congress of Mechanics, Limassol, Cyprus, 12-14 July 2010 IFRC (International Federation of Red Cross and Red Crescent Societies) (2001). World Disaster Report, Focus on Reducing Risk. Geneva, Switzerland, 239 p. IFRC (International Federation of Red Cross and Red Crescent Societies) (2004). World Disaster Report. ISDR (International Strategy for Disaster Reduction) (2005). Hyogo Framework for Action 2005-2015, 21 p. Jostad H.P., Fornes P. and Thakur V. (2013). Effect of strain-softening in design of fills in gently inclined areas with soft sensitive clays. 1st Intern. Wksp Landslides in Sensitive Clays. Québec. October 2013. Jostad H.P. and Grimstad G. (2011). Comparison of distribution functions for the nonlocal strain approach. 2nd International Symposium on Computational Geomechanics, Cavtat-Dubrovnik, Kroatia. Kalsnes B.G., Gjelsvik V., jostad H.P., Lacasse S. and Nadim F. 2013. Risk assessment for quick clay slides - the Norwegian practice. 1st Intern. Wksp Landslides in Sensitive Clays. Québec. October 2013 Kvalstad T.J., Andresen L., Forsberg C.F., Berg K., Bryn P. and Wangen M. 2005a. The Storegga slide: evaluation of triggering sources and slide mechanics. Marine and Petroleum Geology 22, 245-256. Kvalstad T.J., Nadim F. Kaynia A.M., Mokkelbost K.M. and Bryn P. 2005b. Soil conditions and slope stability in the Ormen Lange area, Marine and Petroleum Geology 22: 299-310. Lacasse S. and Boisard, P. Recurrent sliding of underwater slope on Cap Lopez in Gabon. 7th International Symposium on Landslides, Trondheim, Norway. 1, pp. 543–548. Lacasse S., Nadim F., Høeg K. and Gregersen O. 2004. Risk Assessment in Geotechnical Engineering: The Importance of Engineering Judgement. The Skempton Conference, Proc. London. 2, 856-867

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Bishop Bishop Lecture lecture Advanced laboratory testing in research and practice Conférence Bishop Les essais en laboratoire avancés dans la recherche et dans l'industrie Jardine R. J. Imperial College London, UK

ABSTRACT: This lecture demonstrates the special capabilities and practical value of Advanced Laboratory Testing, focusing on its application in advancing the understanding and prediction of how driven piles function and perform in sand. Emphasis is placed on integrating laboratory research with analysis and field observations, drawing principally on work by the Author, his colleagues and research group. The laboratory studies include highly instrumented static and cyclic stress-path triaxial experiments, hollow cylinder and ring-shear interface tests and micro-mechanical research. Soil element testing is combined with model studies in large laboratory calibration chambers, full-scale field investigations and numerical simulations to help advance fundamental methods for predicting pile behaviour that have important implications and applications, particularly in offshore engineering. RÉSUMÉ: Cet exposé décrit les possibilités offertes par les essais en laboratoire de pointe, et en particulier sur leurs apports dans la compréhension et la prévision du comportement des pieux battus dans du sable. L'accent est mis sur l’intégration entre les essais en laboratoire et les observations sur le terrain, à partir des travaux de l'Auteur, ses collègues et leur groupe de recherche. Les essais décris incluent des essais triaxiaux statiques et cycliques avec des appareils suréquipés, des essais au triaxial à cylindre creux, des études d'interfaces pieu/sable à l'aide d'appareils de cisaillement annulaire et des recherches sur la micro-mécanique. Les essais en laboratoire sont combinés à des expériences en chambre de calibration, des études « grandeur nature » sur site et des simulations numériques afin d'aider à l'amélioration des méthodes de prévision du comportement des pieux, qui ont des conséquences importantes en pratique, notamment pour l'industrie offshore. KEYWORDS: Sand; laboratory element tests; non-linearity anisotropy breakage time-dependence; driven piles; field and model tests MOTS-CLÉS: Sable ; tests élémentaires en laboratoire; non-linearité, anisotropie, fragmentation; comportement en fonction du temps; pieu battu; pieu foncé; tests sur le terrain 1

INTRODUCTION

The Bishop Lecture was inaugurated by Technical Committee TC-101 (formerly TC-29) of the ISSMGE, honouring the legacy of Professor Alan Bishop (1920-1988), the leading figure of his generation in geotechnical laboratory experiments and equipment design. Bishop was well known for his meticulous attention to detail, analytical rigour and application of fundamental research in civil engineering practice. His contributions to soil sampling and testing were summarised in the last major keynote he gave, at the Stockholm ICSMFE; Bishop 1981. Similarly admirable attributes were clear in the first Bishop Lecture presented by Tatsuoka 2011, making the invitation to deliver the 2nd Lecture both a considerable challenge and a poignant honour for this former student of Bishop and Skempton. The lives, work and archived papers of the latter two pioneers are described together in a website hosted by Imperial College: www.cv.ic.ac.uk/SkemArchive/index.htm. Our key aim is to demonstrate the special capabilities and practical value of Advanced Laboratory Testing, mirroring Bishop’s work and TC-101’s intent in the International Symposia (IS) it convened in Hokkaido 1994, London 1997, Torino 1999, Lyon 2003, Atlanta 2008 and Seoul 2011. We focus on the mechanics of piles driven in sand, a practical problem that was thought fully resistant to ‘theoretical refinement’ by Terzaghi and Peck 1967. The illustration draws principally on work by the Author, his colleagues and research group. In keeping with Bishop’s approach, emphasis is placed on integrating laboratory research, analysis and field observation. The selected topic is significant industrially. Pile stiffness, capacity, cyclic response and long-term behaviour can be critically important to, for example, wind-turbine foundations. However, the key geomechanics issues are complex and cannot

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be addressed fully or reliably with currently available conventional design tools. Database studies and prediction competitions have quantified the significant biases and scatters associated with conventional practice. The Coefficients of Variation (CoV) established by contrasting axial capacity predictions with field tests typically fall around 0.5 to 0.7. Some methods’ predictions scatter around half the measurements while others tend to double the test values (Briaud and Tucker 1988). The capacity CoVs can be halved and biases largely eliminated by applying modern ‘offshore’ methods (Jardine et al 2005b, Lehane et al 2005). But displacement predictions remain unreliable under axial, lateral or moment loads. It is also unclear how cyclic or extended loading should be considered: Kallehave et al 2012, Jardine et al 2012. Improving understanding and predictive ability will benefit a broad range of applications, especially in offshore energy developments. The Author’s research with displacement piles in sand started with highly instrumented field model piles at Labenne (SW France, Lehane et al 1993) and Dunkerque (N France, Chow 1997), where full-scale testing followed. We review some of the full-scale test results below before considering new research prompted by some surprising and significant results. The Dunkerque profile comprises medium-dense fine-tomedium clean silica Holocene marine sand overlain by hydraulic sand fill. Jardine et al 2006, Jardine and Standing 2012 and Rimoy et al 2013 give details of the geotechnical profiles, pile driving records and testing methods. Static and cyclic axial loading tests were conducted on multiple piles, including six 19.3m long 457mm outside diameter driven steel pipe-piles: R1 to R6. Static axial testing involved a Maintained-Load (ML) procedure where load (Q) was applied initially in 200 kN steps that reduced as the tests progressed. Loads were held constant

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

until creep rates slowed to pre-set limits; the piles took between several hours and 1.5 days to reach failure. More rapid ML tension tests that achieved failure with an hour were also conducted after cyclic loading experiments. Testing rate was found to affect displacements but have little influence on shaft capacity. The cyclic tests were controlled to deliver approximately sine-wave load variations at ≈ 1 cycle/minute. The static testing investigated, among other factors, the effects of pile age after driving. Figure 1 presents tension tests on three identical piles that were aged for 9 to 235 days before being failed for the first time. We note:  

The load displacement (Q – δ) curves are practically identical up to Q ≈ 1 MN but then diverge to show marked increases in Qult (the ultimate load shaft capacity) with age. Creep displacements (dδ/dt when dQ/dt = 0) were negligible until Q > 1 MN after which creep became progressively more important, finally dominating as failure approached.

Load-displacement behaviour was highly non-linear. The overall pile head secant stiffnesses k = Q/δ all fell as loading continued with no discernible ‘linear-elastic’ plateau. This feature is highlighted in Fig. 2 with data from ‘1st time’ tension tests on five ‘R’ piles. The pile stiffnesses, kl, are normalised by kRef, the value developed under QRef - the first (200 kN) load step. The loads Q are normalised by QRef.

An objective assessment was made of how well the Dunkerque pile tests could be predicted by well-qualified engineers by inviting entries to an open competition that concentrated on the static and cyclic tests conducted ≈ 80 days after driving; Jardine et al 2001a. Over 30 (many prominent) international practitioners and academics took part, sending in a wide spread of predictions. The axial capacity estimates confirmed the expected CoV of 0.6, as well as significant bias; the stiffness predictions were similarly spread. No competitor was prepared to predict the cyclic test outcomes; some indicated that cycling should have no effect in clean sand. Figure 3 illustrates the field outcomes in a cyclic failure interaction diagram. The conditions under which 13 tests ended in failure and one developed a fully stable response are summarised by plotting the normalised cyclic load amplitude Qcyc/Qmax static against the average mid-cycle load Qmean/Qmax static where Qmax static = QT current tension capacity. If cycling and testing rate had no effect, then failures should lie on the ‘top-left to bottom-right’ diagonal static capacity line: Qcyc + Qmean = QT in Fig. 3. However, the cyclic test failure points all fell well below this limit, proving a negative impact that grew directly with Qcyc/Qmean. High-level two-way (tension and compression) cycling could halve shaft capacity within a few tens of cycles. Rimoy et al 2013 discuss the piles’ permanent displacement and cyclic stiffness trends, noting also that their non-linear cyclic stiffnesses depended primarily on Qcyc/QT and did not vary greatly with the number of cycles (N) until failure approached. The permanent displacement trends were more complex, depending also on Qmean/QT and N. Interactions were seen between the piles’ ageing and cyclic behaviours: low-level cycling accelerated capacity growth while high-level cycling slowed or reversed the beneficial capacity trend.

Fig. 1. Load-displacement curves from first-time tension failures on Dunkerque piles R1, R2 and R6: Jardine et al 2006

1.0

0.8 Fig. 3. Axial cyclic interaction diagram for full–scale cyclic tests on piles driven at Dunkerque: Jardine & Standing 2012

0.6 kl/kRef

We consider below eight research themes that addressed the shortfalls in understanding revealed by the Dunkerque tests:

0.4

0.2

0.0 0

R2 - R6

5

10

15

20

Q/QRef Fig. 2. Stiffness load-factor curves from 1st time tests at Dunkerque conducted (except R6) around 80 days after driving: Rimoy et al 2013

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1. Characterising the sands’ true stress-strain relationships, correlating advanced laboratory and in-situ measurements. 2. Checking, through Finite Element (FE) modelling, whether laboratory-based non-linear predictive approaches led to better matches with full scale behaviour. 3. Stress-path laboratory testing programmes that investigated creep and ageing trends. 4. Studying the stress conditions imposed by pile installation through highly instrumented Calibration Chamber tests. 5. Grain-crushing and interface-shear zone studies involving high pressure triaxial, ring-shear and laser particle analysis. 6. Quantitative checking against advanced numerical analyses.

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7. Model-pile Calibration Chamber cyclic loading experiments. 8. Cyclic soil element tests to replicate pile loading conditions.

Percentage fine by weight (%)

100

A common theme is that sands show strong non-linearity, plasticity and time dependency from very small strains and have markedly anisotropic properties. It is argued that their overall responses can be understood within a critical state soil mechanics framework, provided that the above features are accommodated and the importance of particle breakage is recognised, especially under high pressures and within abrading shear bands. Space constraints limit the details that can be reported for the various studies cited, or the reviews that can be made of research by other groups. However, PhD theses and coauthored articles are cited to cover the main omissions.

80

60

40 Dunkerque, Kuwano(1999) new-HRS Kuwano (1999) NE34, Yang et al. (2010) TVS, Rimoy & Jardine (2011)

20

0 0.01

0.1

1

10

Particle size (mm)

2

Fig. 4. Summary of particle size distributions for granular media employed in reported laboratory research

CHARACTERISING STRESS-STRAIN BEHAVIOUR

Bishop recognised at an early stage that geotechnical stressstrain measurements are constrained heavily by equipment capabilities. ISSMGE Technical Committee 29 (now TC-101) was set up to coordinate advanced laboratory developments, leading to a review of apparatus, sensors and testing strategies by Tatsuoka et al 1999. The hydraulic stress path cells and Hollow Cylinder Apparatus (HCA) advocated by Bishop and Wesley 1974 and Bishop 1981 allow in-situ stress conditions to be imposed and studies made of shear strength anisotropy; see for example Hight et al 1983 and Shibuya et al 2003a,b. Burland and Symes 1982 and Jardine et al 1984 went onto show that endbedding, sample tilting and compliance caused very large errors in conventional geotechnical strain measurements that often led to completely misleading soil stiffness characteristics. Local strain sensors or dynamic non-destructive techniques are required to obtain representative data: see Tatsuoka et al 1999. Laboratory research with such equipment that contributed to the first phase of research that advanced the “Dunkerque agenda” included the PhD studies of Porovic 1995, who worked with a Resonant Column (RC) equipped HCA and Kuwano 1999 who developed dual-axis Bender Elements (BE) and enhanced resolution local strain sensors for stress-path triaxial tests. Porovic worked mainly with Ham River Sand (HRS), a silica sand graded from Thames Valley gravels that has been tested since Bishop’s arrival at Imperial College and is now known generically as Thames Valley Sand (TVS); Takahashi and Jardine 2007. Kuwano studied Dunkerque sand, spherical glass ballotini and HRS; Connolly 1998 undertook RC and HCA experiments on Dunkerque sand. The sands were tested saturated after pluviation to the desired initial void ratios; Table 1 and Fig. 4 summarise their index properties. Figures 5 to 7 illustrate the apparatus employed in this first period of ‘sand’ research. We consider studies with the Thames Valley (TVS) and French Fontainebleau NE34 sands later in the paper. Table 1. Index properties of silica sands employed in laboratory studies.

Sand

d10 Specific gravity (Gs) (mm)

d50 (mm)

d90 (mm)

Cu

emax

emin

Dunkerque

2.65

0.188

0.276

0.426

2.27

0.97

0.51

NE34

2.65

0.150

0.210

0.230

1.53

0.90

0.51

HRS

2.66

0.190

0.283

0.312

1.64

0.85

0.55

TVS

2.66

0.160

0.250

0.265

1.67

0.85

0.55

Fig. 5. Automated hydraulic stress path triaxial cell for 100mm OD specimens employed to investigate non-linear, anisotropic, pressure and time-dependent stiffness of sands: Kuwano and Jardine 1998, 2002a

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Fig. 6. Bender element configuration to investigate stiffness of sands: Kuwano and Jardine 1998, 2002a

Displacement Transducer

Bellofram cylinder

Ram Clamp Sprocket and torque transmission chain Stepper motor for torsion

Rotary tension cylinder

Hardin oscillator

Tie rod

Proximity transducers

Cam Acrylic chamber wall

Specimen

(Y1) true yield surface that is dragged with the current effective stress point, growing and shrinking with p΄ and changing in shape with proximity to the outer, Y3 surface; Jardine 1992. The latter corresponds to the yield surface recognised in classical critical state soil mechanics. Behaviour within the true Y1 yield surface is highly anisotropic, following patterns that evolve if K, the ratio of the radial to vertical effective stress (K = σ΄r /σ΄z), changes. Plastic straining commences once the Y1 surface is engaged and becomes progressively more important as straining continues along any monotonic path. An intermediate kinematic Y2 surface was identified that marks: (i) potential changes in strain increment directions, (ii) the onset of marked strain-rate or time dependency and (iii) a threshold condition in cyclic tests (as noted by Vucetic 1994) beyond which permanent strains (or p΄ reductions in constant volume tests) accumulate significantly. The Y3 surface is generally anisotropic. For example, the marked undrained shear strength anisotropy of sands has been identified in earlier HCA studies (Menkiti 1995, Porovic 1995, Shibuya et al 2003a,b) on HRS. The surface can be difficult to define under drained conditions where volumetric strains dominate. Kuwano and Jardine 2007 suggested that its evolution could be mapped by tracking the incremental ratios of plastic to total strains. They also suggested that the Phase Transformation process (identified by Ishihara et al 1975, in which specimens that are already yielding under shear in a contractant style could switch abruptly to follow a dilatant pattern) could be considered as a further (Y4) stage of progressive yielding. Jardine et al 2001b argue that the above in-elastic features can be explained by micro-mechanical grain contact yielding/slipping and force chain buckling processes. The breakage of grains, which becomes important under high pressures, has also been referred to as yielding: see Muir-Wood 2008 or Bandini and Coop 2011. HCA testing is necessary to investigate stiffness anisotropy post-Y1 yielding; Zdravkovic and Jardine 1997. However, crossanisotropic elastic parameter sets can be obtained within Y1 by assuming rate independence and combining very small-strain axial and radial stress probing experiments with multi-axis shear wave measurements. Kuwano 1999 undertook hundreds of such tests under a wide range of stress conditions, confirming the elastic stiffness Equations 1 to 5. Ageing periods were imposed in all tests before making any change in stress path direction to ensure that residual creep rates reduced to low proportions (typically 0) stages fall from 0.30 to 0.23 as loading continues, indicating an increasingly plastic response. However, the additional plastic strains developed during creep stages (where dp΄/dt = dεe/dεep = 0) become progressively more significant as loading continued and contributed the major part of the overall ‘consolidation’ strains (εcon) by the end of the test. The latter point is emphasised in Fig. 13 by plotting the proportion of the overall consolidation strain εcon that was due to creep εcre during the pause periods of test H4 and two otherwise identical experiments on loose HRS and medium-dense, nearly spherical, GB. Overall, the relative contribution of creep appears to (i) grow with stress level and grain angularity and (ii) fall with initial void ratio, OCR and stress ratio K = σ΄3/σ΄1. Jardine and Kuwano 2002a also show that creep strain rates decay inversely with time over the first few hours. Jardine et al 2001b offer observations on the micro-mechanical processes that control the experimental behaviour seen in triaxial and HCA tests.

Fig. 11. Effective stress paths followed in drained ‘Creep’ stress path tests on HRS and GB specimens: Kuwano and Jardine 2002a

Fig. 12. Overall e-p΄ relationship of K0 compression tests on mediumdense HRS, showing ratios dεe/dεep of elastic to plastic strains and timedependent compression over creep stages (C): Jardine et al 2001b.

Kuwano and Jardine illustrated aspects of short-term creep behaviour through tests on saturated Ham River Sand (HRS) and Glass Ballotini (GB) specimens prepared at various initial densities. The tests advanced along the drained ‘near isotropic’ and ‘K0’ stress paths set out in Fig. 11 at mean stress rates dp΄/dt of around 100 kPa per hour. The paths were punctuated, as indicated, by periods ‘C’ where samples were allowed to creep under constant stresses for several hours.

It is argued later that the kinematic conditions applying close to the shafts of displacement piles impose approximately constant volume conditions. The constant volume creep response is illustrated in Fig. 14 by showing first the effective stress path followed by an isotropically normally consolidated mediumdense HRS specimen that was allowed to creep to a stable condition before being sheared undrained in triaxial compression

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

medium-dense TVS sand (see Fig. 4 and Table 1) in the advanced hydraulic stress path cell system illustrated in Fig 16.

under a constant axial rate of 0.5%/hour, punctuated by seven constant stress creep pauses. Figure 15 presents the strain-time (ε – t) responses observed over the undrained creep stages. Note: (i) very little creep before the Y2 surface is engaged (at q ≈ 30 kPa ≈ 0.15p΄) (ii) the post Y2 family of ε – t curves in which creep rates grow exponentially with q (iii) a marked softening of the stress-strain response and anti-clockwise effective stress path rotation at the Y3 stage (when q ≈ 160 kPa), (iv) the Y4 Phase Transformation Point (at q ≈ 200 kPa, p΄ ≈ 170 kPa when q/p΄ approaches Mcritical state) and (v) a second family of ε – t curves applying post Y4 showing creep rates that grow slowly as q increases very significantly.

Fig. 15. Strain-time paths followed in seven undrained ‘Creep stages’ of stress-path test H2 on HRS specimen indentified in Fig. 14: Kuwano and Jardine 2002a

Fig. 13. Ratios of creep strains εcre to total consolidation axial strains εcon in K0 compression tests on HRS and GB specimens following paths shown in Fig. 11: Kuwano and Jardine 2002a

p΄ (kPa) Fig. 14. Effective stress paths followed in undrained ‘Creep’ stress-path test H2 on HRS specimen: Kuwano and Jardine 2002a

The triaxial trends bear out the pile load-test trends in Fig. 1 for ‘creep-yielding’ (noted at Q ≈ 1 MN with the R piles) followed by creep rates that rise rapidly with each subsequent load step. It is clear that time-dependency has an important impact on both laboratory and field pre-failure behaviour. We consider next longer-term triaxial stress path experiments designed to investigate the interactions between pile ageing and low-level cyclic loading noted by Jardine et al 2006. Rimoy and Jardine 2011 report suites of tests conducted on

Fig. 16. Advanced IC automated hydraulic stress-path triaxial apparatus and instrumentation for 100mm OD specimens described by Gasparre et al 2007 and employed by Rimoy and Jardine 2011

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and increases in K0. Bowman and Soga (2005) noted similar features in independent experiments, speculating that this feature might play a significant role in pile capacity growth with age. Rimoy and Jardine 2012 also explored interactions between creep and low-level cyclic loading. Figure 20 plots the εs - t trends from tests where the deviator stresses q were varied by one cycle per minute (as in the Dunkerque pile tests) while keeping p΄ constant. The cycling commenced as soon as the stress path arrived at the desired p΄ level with (half peak-totrough) amplitudes qcyc equal to 5, 10 and 15% of p΄. The cyclic tests showed augmented rates of permanent strain development, which in the qcyc = 0.15p΄ test doubled those seen in the ‘true creep’ experiment. Other experiments showed that prior drained ageing (creep) or overconsolidation slow permanent strain development.

1000

q (kPa)

800

600

True creep or cyclic loading with constant p'

1.33

CSL 400

0.868 200

True creep Cyclic loading with constant p'

Ko line

0.20 Creep, p' = 600kPa

0 200

400 600 p' (kPa)

800

0.16

1000

Fig. 17. Effective stress paths followed in creep-cyclic interaction stresspath triaxial tests on TVS specimens: Rimoy and Jardine 2011

Creep, p' = 400kPa

V o lu m etric strain s (% )

0

Creep, p' = 200kPa

0.12 0.08

Figure 17 sets out the effective stress paths followed by Rimoy and Jardine 2011, indicating the pause points at which drained creep straining was observed for 2 to 4 day durations under constant stresses - either in an undisturbed ‘true’ state or in combination with low-level drained cyclic loading.

0.04 0.00

-0.04 0

0.20% 0.18%

Creep, p' = 400kPa

0.16%

2000

3000 minutes

4000

5000

6000

Fig. 19. Volume strain-time trends followed in ‘true creep’ stages of stress-path triaxial tests on TVS specimens: Rimoy and Jardine 2011

Creep, p' = 200kPa

0.14%

0.30

0.12%

qcyc, 0.05p' = 30kPa

0.10%

qcyc, 0.025p' = 15kPa

0.25

0.08%

qcyc, 0.015p' = 10kPa

0.06% 0.04% 0.02% 0.00% 0

1000

2000

3000 minutes

4000

5000

6000

Fig. 18. Shear strain invariant-time trends followed in ‘true creep’ stages of stress-path triaxial tests on TVS specimens: Rimoy and Jardine 2011

ε c y c a x ia l - ε c re ep (% )

Shear strain invariant (% )

1000

Creep, p' = 600kPa

0.20 0.15 0.10 0.05 0.00

Figures 18 and 19 show the volumetric and shear strain invariant responses observed during ‘true’ creep at three p΄ levels, showing stable and consistent trends. While the invariant shear strain increased monotonically with time and p΄ level, the volumetric trends reversed when εs exceeded ≈ 0.015% after several hours and diverged strongly from the initially near K0 pattern, where dεa/dεvol = 1 and dεs/dεvol = 2/3 for zero radial strains. Monotonically continuing shear distortion led to sharp rotation of strain increment directions, eventually establishing a steady trend for dεs/dεvol ≈ -1. This interesting kinematic yielding trend, which was not apparent in the shorter duration creep tests investigated by Kuwano 1999, can be seen as the (stationary) effective stress point engaging a kinematic yield surface that is moving with respect to time or strain rate. Given the final strain increment direction, it appears that the Y2 ‘bubble’ has moved rightwards with time and the fixed effective stress point has engaged its leftward limit. Under strain-controlled K0 conditions any radial dilation has to be suppressed, leading to radial effective stresses

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0

1000

2000

3000 Cycles

4000

5000

6000

Fig. 20. Shear strain invariant-time trends from cyclic stress-path tests on TVS specimens conducted at 1cycle/minute: Rimoy and Jardine 2011

More complex interactions are revealed by plotting εs against εvol in Fig. 21. It can be seen that cyclic loading retards the shift from contractive-to-dilative volumetric response. The timedependent Y2 point is pushed forward in terms of both creep duration and shear strain developed. Low-level cyclic loading does not simply accelerate creep. It also holds back and probably expands the time-dependent kinematic Y2 surface. It is interesting that low-level cycling enhances pile capacity growth, suggesting that the delayed dilation mechanism may be playing a more complex role than had been appreciated in pile axial capacity growth with time. The laboratory tests provide critical data against which new time-dependent and kinematic yielding models may be tested.

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soil sensors. Zhu et al 2009 focus on the sensors’ calibrations and performance, emphasizing the care needed to address nonlinear and hysteretic cell action.

0.35 qcyc/p' = 0.05 p'=600kPa

Shear strains invariant (%)

0.30

qcyc/p' = 0.025 p'=600kPa qcyc/p' = 0.015 p'=600kPa

0.25 Pure creep at p' = 600kPa Pure creep at p' = 400kPa

0.20

Pure creep at p' = 200kPa

0.15 Ko line

0.10 0.05 Yield points

0.00 0.00

0.05

0.10 0.15 0.20 0.25 Volumetric strains (%)

0.30

0.35

Fig. 21. Shear strain invariant-volume strain trends followed in creepcyclic interaction stress-path triaxial tests on TVS specimens: Rimoy and Jardine 2011

ESTABLISHING THE DEVELOPED AROUND DISPLACEMENT PILES

STRESS CONDITIONS LABORATORY MODEL

The laboratory element testing described above reveals highly non-linear, anisotropic, time-dependent and in-elastic stressstrain behaviour. These features depend critically on the samples’ effective stress states and stress histories. However, the lack of knowledge regarding the effective stress regime set up in the surrounding sand mass when piles are driven called for further research. Calibration Chamber experiments offered the promise of new insights that would help to link laboratory element tests and field pile behaviour. Laboratory Calibration Chambers (CC) were developed originally to aid field SPT and CPT interpretation in sands. Multiple test series have been conducted on uniform (wellcharacterized) sand masses under controlled pressure or displacement boundary conditions; see for example Baldi et al 1986 or Huang and Hsu 2005. Laboratory CCs also provide scope for measuring stresses in soil masses around model piles (during and after installation) and also allow ‘post-mortem’ sand sampling; these activities are far more difficult to perform in field tests. Joint research with Professor Foray’s group at the Institut National Polytechnique de Grenoble (INPG) has included a comprehensive study of the stresses developed around closedended displacement piles. Cone-ended ‘Mini-ICP’ stainlesssteel, moderately rough (RCLA ≈ 3μm) piles with 18mm radii R (the same as a standard CPT probe) were penetrated 1m into dry, pressurized, and highly instrumented medium-dense Fontainebleau NE 34 silica sand. NE 34 has the index properties shown in Fig. 4 and Table 1 and is broadly comparable to the earlier discussed Dunkerque, HRS and TVS sands. Jardine et al 2009 detail the general experimental arrangements outlined in Fig. 22. Cyclic jacking, with full unloading between strokes, was imposed to simulate pile driving installation. The Mini-ICP instrumentation included reduced-scale Surface Stress Transducers that measure radial and shear shaft stresses at radial distances r/R = 1 from the pile axis at three levels, as shown on Fig. 23. Measurements were also made of σ΄z, σ΄θ and σ΄r at two to three levels in the sand mass at radial distances between 2 and 20R from the pile axis using miniature

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Fig. 22. Schematic arrangements for fully instrumented environmentally controlled Calibration Chamber Mini-ICP tests: Jardine et al. 2009 10 1500 1400 1300 1200

Distance from pile tip, h (mm)

5

d

1100

Axial load

1000

Surface stress transducer

900 800

Trailing cluster 700 600 500

Following cluster

400 300 200 100 0

 1

Leading cluster and Pile tip

 1

Fig. 23. Schematic of laboratory Mini-ICP pile with three levels of Surface Stress Transducers, as well as Axial Load Cells, temperature sensors and inclinometers: Jardine et al 2009 Upper annular membranes were used to apply a surcharge pressure of σ΄zo ≈ 150 kPa to the sand mass. Separate CPT tests established qc profiles for various boundary conditions. As shown in Fig. 24, two alternative membrane designs gave quasi-

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

constant CPT trace sections with qc = 21±2 MPa, although this was achieved at a shallower depth with the smaller Internal Diameter (ID) membrane. Also shown is the qc profile predicted by Zhang et al 2013 that is discussed later. Rimoy 2013 describes more recent experiments with the same equipment, noting that axial capacities from multiple load tests agree encouragingly well with predictions made with the ‘field-calibrated’ capacity approach outlined by Jardine et al 2005b, which gave good results for the Dunkerque field tests.

penetration (σ΄rm) stages. The results are normalized for local qc and plotted with cylindrical co-ordinates defined relative to the pile tip. Normalised vertical distances (h/R) above are positive, points below have negative h/R. Separate plots were derived for ‘stationary’ pause radial stresses (σ΄rs points) recorded when the pile head was unloaded fully. Moving and stationary contour sets were also reported for the vertical (σ΄z) and hoop (σ΄θ) stresses. 3.0

qc (MPa) 5

10

15

2.5

20

0

Penetration (mm)

h/R=5.6

1.5

h/R=16~21 1.0

h/R=31.1

400 200mm ID top membrane 50mm ID top membrane Numerical simulation

600

0.5

h/R=40.6

0.0 0

5

800

15

20

r /R

1200

Fig. 24. Measured and predicted qc profiles with alternative CC topmembranes: Jardine et al. 2013a and Zhang et al 2013 10

50

0 0

0.25 2.0

1.0

40

1.0 2.0

4.0

3.0

30

10

Fig. 26. Radial profiles of radial stresses measured around model pile after installation in laboratory Calibration Chamber (normalized by qc and shown in %): Jardine et al. 2013b

1000

0.25

5

6.0

4.0

8.0

1.5 5.0

20

1

6.0

h/R

h/R

2.0

'

200

10

0

8.3

6.0

4.0

10

2.0

3.0

The contour plots indicate intense stress concentrations emanating from the pile tip. Radial stress maxima exceeding 15% qc were observed at h/R~0.5, r/R=2 during penetration, while the ‘zero-load’ stationary values were 2 to 3 times smaller. Yang et al 2010 describe how an active failure develops beneath the advancing tip where, on average, σ΄zm/qc = 1, σ΄rm = σ΄θm = KAσ΄zm and KA = tan2(45 + φ'/2). Close analysis of the ‘moving’ and stationary’ stresses measurements shows the greatest divergence near the tip (-5 0.2mm sands, to lesser δ angles after 50mm displacements than equivalent upper interface tests, where fine fragments can fall from above into void spaces beneath the shear zone. Lower interface ring-shear tests gave similar trends at 50mm displacement to (5mm) direct shear interface tests.

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

Fragments appear to choke available void spaces after large displacements (8m), preventing lower friction angles persisting with coarser sands and upper interfaces. The ring shear trends converge, but do not conform fully to the uniform δ = 29o CUR 2001 recommendation.

The Calibration Chamber model studies reported in Section 5 testified to the extreme stresses developed beneath advancing pile tips. Stresses rose and fell around the shaft (at any given depth) by almost two orders of magnitude as the tip penetrated to greater depths. Such changes in stress level, combined with particle breakage, affect the sand’s constitutive behaviour. Altuhafi and Jardine 2011 conducted tests to investigate these features using the high pressure apparatus shown schematically in Fig. 35 to subject medium-dense NE 34 to the effective stress paths set out in Fig. 36.

shear strength and dilatancy of the overconsolidated’ and partially crushed sand.

‘heavily

See Fig. below for low pressure test stages

Fig. 35. High pressure triaxial apparatus employed to test crushing NE34 sand. System described first by Cuccovillo and Coop 1998

The key test stages were:  





K0 compression to p΄ = 9 MPa, simulating the pile tip advancing towards the sand element from above. Drained compression under constant σ΄r until apparent ‘critical states’ were reached with σ΄1 > 20 MPa, simulating failure beneath the conical pile tip. Tests that stopped abruptly developed large creep strains. The displacement strain rates therefore were slowed progressively to reduce residual creep effects prior to unloading. The ‘critical state’ e-p΄ relationships depend on time. Drained unloading to q = 0 under constant σ΄r before isotropic unloading to p΄ values between 150 and 500 kPa (giving ‘OCRs’ of 40 to 140 in terms of vertical stresses), simulating the sharp unloading experienced as the tip passes. Renewed drained shearing to failure at constant σ΄r in compression (or at constant p΄ in extension) to assess the

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Fig. 36. Effective stress paths followed in high-low pressure triaxial tests on NE 34 sand, showing high pressure stages (top) and overconsolidated low pressure stages (below): Altuhafi and Jardine 2011

The results obtained are illustrated in Fig. 37, plotting mobilised angles of shearing resistance φ΄ against axial strain. The upper plot (a) shows the generally ductile-contractant response seen in six similar high pressure tests, with peak φ΄ only slightly greater than the ‘critical state’ (30o) angle. The lower plot (b) summarises the ‘overconsolidated’ response observed on recompression after unloading. All three ‘overconsolidated’ samples dilated as they sheared, developing peak φ΄ ≈ 42o, well above the ultimate angles (around 33o) developed after large shear strains and diminished dilation. It is clear that the sand’s behaviour alters radically on unloading as the pile tip advances by a few diameters, changing from being contractant, ductile, highly prone to creep and

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offering relatively low φ΄ beneath and around the tip, to being dilatant, brittle and able to mobilise far higher peak φ΄ in the mass that surrounds the shaft. These features were critical to Jardine et al 2013b’s interpretation of the model pile Calibration Chamber stress measurements illustrated above in Figures 24 to 27. Further analysis of the evolving family of ‘critical state’ e-p΄ curves developed by crushing is underway by Dr Altuhafi.

', Degrees

50

40

30 '

Ultimate  =30o

P-T1 P-T2 P-T3 P-EE1 P-EE2 P-EE3

20

10

0 0

10

20

30

40

Strain% Fig. 38. Comparison between (a) Yang et al’s interpretation of breakage around penetrating Mini-ICP model piles and (b) simulation breakage parameter B contours for same tests; Zhang et al 2013

50

' , Degrees

Peak '= 42o 40

6.0

30

Ultimate  '= 33

(a) Numerical results by Zhang et al. (2013) o

Fontainebleau sand

P-T1 P-T2 P-T3

4.5

r / qc: %

20

10

h/R=6

'

0

h/R=3 3.0

0

5

10

15

20

25

Strain%

h/R=9

1.5

Fig. 37. Mobilised φ΄ values plotted against axial strain for both high (a) and low (b) pressure test stages of triaxial tests on NE34 sand: Altuhafi and Jardine 2011

0.0 0

10

15

20

r /R

Fig. 39. Radial profiles of σ΄r/qc from Zhang et al 2013’s analysis of Mini-ICP pile in NE 34 sand

Recently published numerical analyses allow further links to be established between the soil element and model pile experiments. Zhang et al 2013 present FE analyses of penetration in sands in which they adopted an Arbitrary Lagrangian Eulerian (ALE) approach to deal with the implicit moving boundary problem and a constitutive model that accounted for grain size distribution evolving through grain breakage. Their analyses included simulations of the Calibration Chamber (CC) model pile tests that applied a ‘breakage’ constitutive model that they calibrated against NE 34 laboratory tests reported by Yang et al 2010 and others. Zhang et al’s predictions for the Mini-ICPs end-bearing characteristics were presented in Fig. 24, together with the CC measurements. The agreement is good when considering the same CC upper boundary conditions. Figure 38 compares the breakage pattern identified by Yang et al 2010 around the MiniICP pile tip with Zhang et al 2013’s contoured predictions for their internal breakage parameter B, which scales linearly between the sand’s initial (B = 0) and ultimate (B = 1.0) ‘fully crushed’ grading curves. The simulated and experimentally established patterns are similar, with the maximum B predicted as ≈ 0.35 close to the shaft, far from the ‘fully broken’ B = 1 limit. The grading curves’ predictions match Yang et al’s measurements well in all three zones, although they do not recover the experimentally observed Zone 1 thickness growth with pile tip depth h/R. The latter is thought to develop through the un-modelled process of cyclic interface shear abrasion.

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6.0

(a) Numerical results by Einav (2012) Fontainebleau sand h/R=3

4.5

h/R=6 3.0

'

COMPARISON WITH NUMERICAL ANALYSES

 / qc: %

7

5

h/R=9

1.5

0.0 0

5

10

15

20

r /R

Fig. 40. Radial profiles of σ΄θ/qc from Zhang et al 2013’s analysis of Mini-ICP pile in NE 34 sand.

Correspondence with Zhang, Nguyen and Einav led to further processing of the stress predictions implicit in their numerical analyses. Interesting comparisons are presented from Yang et al 2013 in Figs. 39 and 40, plotting the σ΄r and σ΄θ predictions transmitted by Professor Einav against r/R. The stresses are normalised by predicted qc, as are the experimental equivalents shown in Figs. 26 and 27. The overall trends show

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encouraging quantitative agreement when comparisons are made between predictions and measurements made at h/R values up to 10; see for example the match between the common curves given for h/R ≈ 6. Naturally, scope exists to consider further factors such as: the effects of stress history on dilatancy and shear strength; creep behaviour; and the extreme cyclic loading that accompanies pile installation and leads to radial stresses continuing to reduce with h/R at ratios greater than 10.

controlled cycles. The more severe TW test progressed further and developed a full failure system with a ‘butterfly-wing’ effective stress path pattern resulting from slip displacements that generated dilatant loading stages followed by sharply contractant unloading stages. Nf= 1

Nf = number of cycles to failure

Tw

o

LABORATORY MODEL PILE TESTS TO INVESTIGATE CYCLIC LOADING

  

w O ne

Qcyclic/QT

The Mini-ICP Calibration Chamber experiments described in Section 5 included multiple suites of axial cyclic loading tests with the model piles installed into pressurised medium-dense NE 34 sand. Cycling was found to have a broadly similar effect on axial capacity to that seen in the Dunkerque field tests. Figure 41 presents an overall interactive diagram which compares directly with the field patterns in Fig. 3. Tsuha et al 2012 and Rimoy et al 2013 report on the cyclic stiffness and permanent displacement trends. Broadly, they classify responses to cycling as:

5 10

0.4

1 10

4 4 170

100 500

Stable: capacity increasing slightly, displacements small and stabilising) over 1000 or more cycles Unstable: reaching failure with 100 cycles, or Metastable: falling between these limits

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Unstable

0.6

0.2

A particular advantage offered by the laboratory model pile arrangements shown in Figs. 22 and 23 was the ability to measure the pile-sand effective stress path response directly, both at the shaft interface (with the Mini-pile’s leading, Following and Trailing Surface Stress Transducers) and within the sand mass by the sand-stress senor arrays. Figure 42 illustrates the local interface effective stress paths followed under Stable conditions in a 1000 cycle experiment. The patterns resemble those seen in Constant Normal Stiffness (CNS) shear experiments (see for example Boulon & Foray 1986 or Dejong et al 2003) with radial effective stresses increasing under tension loading (that generates negative shaft shear stress) and decreasing under compressive load increments around the relatively rigid Mini-ICPs. While the load-displacement response is in-elastic (non-linear and hysteretic) under even low-level cycling, the radial effective stress changes and pile head movements induced by each cycle are small. The effective stress paths appear to match, approximately, the Y2 criteria described in Section 2 and traced by Kuwano and Jardine 2007 in small strain triaxial probing tests. Rather than remain exactly static, the radial stresses reduced, albeit at very slow rates, over time indicating a tendency towards contraction and migration towards the interface shear failure criterion angles established by Yang et al 2010 through interface ring shear tests, or those shown in Fig. 34 from Ho et al 2011. The continuing rates of radial stress reduction might also be related to very slow rates of continuing interface surface abrasion and particle modification. Multiple static tension tests on the Mini-ICPs showed shaft capacities increasing (by up to 20%) as a result of stable cycling, mainly due to changes in loading stress-path geometry that gave a less contractive response under static loading. The Dunkerque field tests also showed tension capacity increasing after a stable 1000 cycle test; Jardine and Standing 2013. Figures 43 and 44 demonstrate the contrasting responses seen in Metastable tests under One-Way (OW) and Two-Way (TW) loading respectively. All paths approach the interface failure envelope as cycling continues, either asymmetrically under OW loading or more symmetrically in the TW test. The milder OW test shows a similar pattern to the Stable test shown in Fig. 40, except that it migrates more rapidly and engages the critical δ= 27o failure line, leading to the onset of local slip after several hundred load

ay

0.8

66

Meta-Stable

500

1000 >1000

0.0 -0.2

Stable 0.0

0.2

0.4

Qmean/QT

0.6

0.8

1.0

Fig. 41. Effects on shaft capacity of cyclic loading. Interactive stability diagram from Mini-ICP CC tests: Tsuha et al 2012.

Leading A Following B Trailing C

200

Shear stress rz (kPa)

8

w

ay

1.0

'

 =27

o

100

0

-100 Direction of radial stresses

-200 0

100

200

300

400

500

Radial stress 'r (kPa) Fig. 42. Interface shear τrz - σ΄r effective stress paths: Stable cyclic test ICP4-OW1: Tsuha et al 2012. Close examination reveals the top-down progressive failure process described by Jardine 1991, 1994. The points where behaviour switches from contractant to dilatant fall on an interface Phase Transformation line analogous to that noted by Ishihara et al 1975.

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Shear stress rz (kPa)

200

Leading A Following B Trailing C

'

 =27

Tsuha et al 2012 report on the similarly in-elastic cyclic local effective stress responses measured by the multiple cells positioned in the surrounding sand mass, relating these to the sand mass failure criteria established by the experiments outlined in Fig. 37.

o

100

9 0

Predictions can be made through cyclic soil element testing of how cyclic pile head loading affects the local shear stresses rz available on the shaft and shear strains in the surrounding soil; Jardine 1991, 1994. Considering the conditions applying close to axially loaded shafts, as in Fig. 46, the hoop strain  must be zero due to symmetry. Also z must be small if the pile does not slip against the shaft and the pile is relatively stiff. The only significant normal strain components are radial (r) and these are constrained by the radial stiffness of the surrounding sand mass.

-100 Direction of radial stresses

-200 0

100

200

300

400

LABORATORY ELEMENT TESTS TO INVESTIGATE CYCLIC LOADING PROCESSES

500

Radial stress 'r (kPa)

Fig. 43. Interface shear τrz - σ΄r effective stress paths: Metastable cyclic test ICP2-OW3: Tsuha et al 2012.

Leading A Following B Trailing C

Shear stress rz (kPa)

200

'

 =27

o

100

0

-100 Fig. 46. Soil element adjacent to a pile shaft: Sim et al 2013 Direction of radial stresses

-200 0

100

200

300

400

500

Radial stress 'r (kPa) Fig. 44. Interface shear τrz - σ΄r effective stress paths: Metastable becoming Unstable cyclic loading test ICP4-TW1: Tsuha et al 2012

Leading A Following B Trailing C

Shear stress rz (kPa)

200

'

 =27

o

The changes in local radial stress, 'r, developed on the shaft in response to Δrz increments that cause dilative or contractive radial displacementsr at the interface can be related to the shear stiffness of the surrounding sand by the elastic cavity expansion expression given as Eq. 6; Boulon and Foray 1986. Jardine et al. 2005b suggest that r is approximately equal to the peak-to-trough centreline average roughness of the pile surface under static loading to failure. Provided that strains remain very small and the shear stiffness is linear, Eq. 6 implies a Constant Normal Stiffness (CNS) interface shear boundary condition, where KCNS is the interface’s global radial stiffness value. δσ΄r /δr = 2G/R = KCNS

Eq. 6

100

0

-100

Direction of radial stresses

-200 0

100

200

300

400

500

Radial stress 'r (kPa) Fig. 45. Interface shear τrz - σ΄r effective stress paths: Unstable cyclic test ICP2-TW1: Tsuha et al 2012

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Laboratory shear tests can be conducted under CNS conditions (Boulon & Foray, 1986 or Dejong et al 2003) to mimic the pile loading boundary conditions and observe the near-shaft cyclic soil response. Suitable mixed boundary conditions can be devised for simple shear, triaxial or HCA tests. However, sands’ shear stiffnesses are non-linear, pressure dependent and anisotropic. Also KCNS varies with 1/R, making it hard to define meaningful single CNS values. Constant volume tests in simple shear, triaxial or HCA cells provide upper limit, infinite, CNS conditions that can be met by cycling saturated samples under undrained conditions. More sophisticated controls can be imposed if reliable information is available about the interface stress and strain boundary conditions. Constant volume or CNS Simple Shear (SS) tests provide conditions analogous to those near pile shafts; Randolph and Wroth 1981. However, conventional simple shear tests cannot provide a full description of the sample’s stress state: neither

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invariant effective stress paths nor Mohr circles of stress can be drawn. Shen 2013 presents new DEM based simple shear simulations. His analyses, which did not require any assumption of idealised co-axial (or other) plasticity in the sand, emphasize the differences between the true internal stress variables and the ‘average’ stresses deduced from boundary measurements. He also highlights the impact of apparatus details on the parameters interpreted by alternative simple shear failure hypotheses. Shibuya and Hight 1987, Menkiti 1995, Nishimura 2006 and Anh-Minh et al 2011 outline the principles and technicalities of conducting SS tests with HCA equipment. While HCAs are subject to sample curvature effects that have to be considered (Hight et al 1983), their annular geometry automatically provides the complementary shear stresses and so reduces stress nonuniformity. They also allow the full stress and strain tensors to be defined and permit detailed assessments of the effects of anisotropy, variable b values (reflecting σ2 ratios or Lode angles) and principal stress axis rotation. Undrained triaxial experiments can also provide useful information. The shear stress changes Δrz developed on the pile shaft pile and changes to triaxial deviator stress Δq = Δ(1- 3) can be inter-related by assuming an isotropic soil response and applying general stress invariants, or by simply noting that in a Mohr circle analysis increments of pure shear shaft loading Δrz have an equivalent effect to an increment Δq that is numerically twice as large. In this simplified view, the changes to mean effective stress, Δp' observed under cyclic loading in the triaxial cell can be seen as implying approximately equivalent proportional Δ'r changes at points close to the shaft. Sim et al 2013 emphasize the need for very stable high resolution test equipment and stable environments for such tests. This applies particularly to long duration, low-level cycling tests where p΄ drift rates and changes in cyclic stiffness/permanent strain development may be slow. Sim et al also report cyclic experiments on Dunkerque and NE 34 sands designed to help interpret the field and laboratory CC model pile tests. Their ongoing research programme is investigating:    

interaction diagrams (such as those in Figs 3 or 41). If further analysis is warranted, laboratory or field test data can be applied in site-specific and storm-specific calculations that follow either a local (T-Z, the left hand path in Fig. 48) or a global (the right hand route in Fig. 48) assessments procedure. The global approach is most applicable when soil conditions are relatively uniform and progressive top-down failure is not a major concern.

Fig. 47. Leftward migration of effective stress paths over 1500 undrained qcyclic = 0.2 p΄ cycles. Triaxial tests on Dunkerque and NE 34 sands from p΄0 = 150 kPa, OCR = 4: Sim et al 2013

Differences between HCA SS and triaxial responses. Effects of pile installation stress history, including the ‘overconsolidation’ that takes place as the tip passes and the effects of the shearing cycles imposed by jacking or driving. The sequence in which different cyclic load packets are applied, assessing the applicability of Miner’s rule. Varying sand types and initial sand states.

Figure 47 illustrates the leftward effective stress path drifts developed in undrained cyclic triaxial tests with paired tests on medium-dense Fontainebleu and Dunkerque samples conducted after K0 consolidation to 800 kPa and unloading to OCR = 4, to simulate pile installation for points positioned 2 < r/R < 3 from a pile shaft. 1500 qcyclic = 0.20p΄ stress controlled cycles were then applied at 1/per minute. The stress paths evidently engaged the samples’ Y2 surfaces. Slow migration led to final mean effective stress reductions of 30 and 40% overall for NE34 and Dunkerque samples respectively under the stringent constant volume conditions imposed. It is interesting that the effective stress paths remained within the Mini-ICPs τ/σ΄n < tan δ΄ interface shear envelope (δ΄ = 27o when shearing against NE 34 or Dunkerque sand, see Figs. 34 and 42-45) implying that while shaft failure would not be expected to reduce in an equivalent cyclic pile test, the pile shaft would not fail within 1500 cycles. Jardine et al 2005b and 2012 offer guidance on how to apply such laboratory testing to estimate the axial response of offshore piles under storm cyclic loading. Referring to the flow chart given in Fig. 48, the first essential step is careful characterisation (applying rainfall analysis methods) of the storm loads to establish equivalent batches of uniform cycles. Initial screening checks are then recommended with experimentally derived (or appropriately validated theoretical) published cyclic failure

Fig. 48 Flow chart outlining approaches for assessing cyclic loading effects in driven pile design: after Jardine et al 2012. Jardine et al 2012 describe several approaches for such calculations. These include the simple ‘ABC’ formulation given by Jardine et al 2005b. Calibration of the latter approach against both laboratory tests and the Dunkerque field experiments indicated encouraging agreement; Jardine and Standing 2013.

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Recent practical applications include a fleet of 40 wind-turbines at Borkum West II (German North Sea) which employ the tripod design shown in Fig. 49 and each rely on three 2.48m diameter piles driven in (mainly) very dense sands; Merritt et al 2012. Another application of the laboratory derived ‘ABC’ approach involved manned oil platforms founded on pile groups driven in very hard sandy glacial tills: Jardine et al 2012.

3. Behaviour can only be considered elastic within a very limited kinematic true (Y1) yield surface that is dragged with the current effective stress point, growing and shrinking with the mean effective stress p΄ and changing in shape with proximity to the outer, Y3 surface; stiffness is anisotropic within Y1, following patterns that evolve with K = σ΄r /σ΄z. 4. Plastic straining commences once Y1 is engaged and becomes progressively more important straining continues along any monotonic path. 5. An intermediate Y2 kinematic surface may be identified in either continuum or interface shear tests that signifies: (i) potentially marked changes in strain increment directions (ii) the onset of important strain-rate or time dependency and (iii) a threshold beyond which permanent strains (and mean effective stress reductions in constant volume tests) accumulate significantly in cyclic tests. 6. Creep tests and experiments that combine drained creep and low level cycling show that the Y2 process is both time dependent and affected by cyclic perturbations. 7. Undrained cyclic tests taken to large numbers of cycles tend to show continuous rates of p΄ reduction, even under relatively small strain cycles. These trends may be modified considerably by overconsolidation, ageing or pre-cycling. 8. Particle breakage develops under large displacement interface shearing as well as high pressure compression and triaxial conditions. Breakage leads to continuous evolution of the index properties and critical state e-p΄ relationships. Conclusions regarding piles driven in sand include:

Fig. 49. Wind-turbine tripods in fabrication yard; http://www.powertechnology.com/projects/borkum-farm/borkum-farm3.html

The fully analytical cyclic assessment route shown as the central path through Fig. 48 may also be followed. Laboratory testing can provide the detailed information required for modelling the sands’ complex behaviour including: stiffness and shear strength anisotropy; non-linearity and progressive yielding; grain crushing; time effects/creep; and cyclic loading responses. Similarly, the laboratory and field model pile stress measurements can guide the specification (or modelling) of the effective stress regime set up around the driven piles and show how this may change under static/cyclic loading conditions. The stage is now set for numerical modelling that can capture field behaviour far more accurately than was previously possible. 10 SUMMARY AND CONCLUSIONS The key aim of the lecture was to demonstrate the special capabilities and practical value of the Advanced Laboratory Testing promoted by Bishop and TC-101. New insights have been offered through static and cyclic experiments with the apparatus and techniques they advocated, including highly instrumented stress-path and high pressure triaxial tests as well as hollow cylinder, ring-shear interface and micro-mechanical experiments. Emphasis has been placed also on integrating laboratory research, field observations, numerical analysis and calibration chamber model pile studies to advance understanding and prediction of the complex behaviour of driven piles in sands. The experiments investigated sand behaviour under a wide range of conditions. Aspects highlighted for consideration in ongoing and future constitutive modelling include: 1. The strong non-linearity, marked in-elasticity and time dependency seen from small-to-large strains. 2. Markedly anisotropic behaviour within the large scale classical critical state soil mechanics (Y3) yield surface. Sands also show Phase Transformation (Y4) over a wide range of states. These features may occur in either soil continua, or during shearing against interfaces.

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1. Conventional approaches for capacity and load-displacement assessment have generally poor accuracy and reliability. 2. It is possible to improve predictions considerably through numerical analyses that capture the observations made with advanced laboratory stress-strain and interface shear tests. 3. Such predictions rely critically on assumptions regarding the stresses set up around the piles during and after installation. 4. Laboratory and field tests highlight the importance of plastic and time-dependent straining which becomes progressively more important as stress and strain levels rise. 5. The Calibration Chamber model pile tests demonstrate key physical features of the pile-soil mechanics, including the extreme stress changes and grain breakage experienced during installation. Micro-mechanical laboratory analysis and high pressure triaxial and ring shear tests allow the properties of the modified material to be studied in detail. 6. Laboratory model pile experiments demonstrate that radial stress maxima develop at some distance from the pile shafts. This feature can also be predicted analytically in studies that address grain breakage. Taken together with the creep trends discussed above, this feature offers a mechanism for the growth in shaft capacity of piles driven in sand over time. 7. Axial cyclic pile tests show broadly similar modes of Stable, Metastable and Unstable behaviour in full scale field tests and model experiments in Calibration Chambers. 8. Local stress measurements made on the ICP and Mini-ICP piles give profound insights into the mechanisms of cyclic degradation, demonstrating features of kinematic yielding and interface shear failure that can be tracked in triaxial, HCA and ring shear laboratory experiments. Advanced laboratory testing is critical to advancing all difficult geotechnical engineering problems where the outcomes depend critically on the detailed constitutive behaviour of the ground. Tatsuoka 2011, for example, described advanced testing directed towards the performance of large bridge foundations and the compaction of reinforced earth retaining wall backfills, while Kovacevic et al 2012 describe novel analyses of very large submarine slope failures that employed models derived also from detailed and advanced laboratory studies.

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12 ACKNOWLEDGEMENTS The Author acknowledges gratefully the many contributions by current and former co-authors, students, technicians, colleagues and co-workers principally at Imperial College, but also at: Building Research Establishment (BRE, UK), Cambridge-Insitu (UK), D’Appolonia (Italy), Geotechnical Consulting Group (GCG, London), IFP (France), INPG (Grenoble, France) and ISSMGE TC-29/101. He also acknowledges with thanks funding from the Commonwealth Commission, CNRS (France), EPSRC (UK), EU, HSE (UK), NSFC (China), Royal Society (UK), Shell (UK), Total (France) and other bodies. Prof. David Hight and Dr Jamie Standing are thanked also for their useful comments on the manuscript. 13 REFERENCES Addenbrooke, T.I., Potts, D.M. and Puzrin, A.M. 1997. The influence of pre-failure stiffness on the numerical analysis of tunnel construction. Géotechnique, Vol 47, No 3, pp 693712. Altuhafi, F. and Jardine, R.J. 2011. Effect of particle breakage and strain path reversal on the properties of sands located near to driven piles. Deformation Geomaterials. Proc. ISSeoul, Hanrimwon, Vol. 1: 386-395. Anh-Minh, N., Nishimura, S., Takahashi, A. and Jardine, R.J. 2011. On the control systems and instrumentation required to investigate the anisotropy of stiff clays and mudrocks through Hollow Cylinder Tests. Deformation Characteristics of Geomaterials. Proc. IS-Seoul, Hanrimwon, Vol. 1: 287294. Baldi, G., Bellotti, R., Ghionna, V., Jamiolkowski, M. & Pasqualini, E. 1986. Interpretations of CPTs and CPTUs, 2nd part: Drained penetration of sands. 4th Int Conf on field instrumentation and in-situ measurements, Singapore: 143156 Bandini, V. and Coop, M.R. 2011. The influence of particle breakage on the location of the critical state line of sands. Soils & Foundations, 51 (4): 591-600 Bishop, A.W., Green, G.E., Garga V.K., Andresen, A. and Brown, J.D. 1971. A new ring shear apparatus and its application to the measurement of residual strength. Géotechnique, 21 (4): 273-328. Bishop, A.W. and Wesley, L.D. 1974. A hydraulic triaxial apparatus for controlled stress path testing. Géotechnique, 25 (4): 657-670. Bishop, A.W. 1981. Thirty five years of soil testing. Proc 10th ICSMFE, Stockholm, LiberTryck, Vol. 4: 185-195. Boulon, M. and Foray, P. 1986. Physical and numerical simulation of lateral shaft frictions along offshore piles in sand. Proc. 3rd Int. Conf. on Numerical methods in Offshore Piling, Nantes: 127 - 147. Bowman, E.T. and Soga, K. 2005. Mechanisms of set-up of displacement piles in sand: laboratory creep tests. Canadian Geotechnical Journal, 42 (5): 1391-1407. Briaud J.L. and Tucker, L.M. 1988. Measured and Predicted Axial Response of 98 Piles. ASCE Journ. Geot. Engrg. Vol 114, No. 9, pp 984-1001. Burland, J.B. and Symes, M. 1982 A simple axial displacement gauge for use in the triaxial apparatus. Géotechnique 32, 1, pp 62-65. Burland, J.B. and Burbridge, M.C. 1984. Settlement of foundations on sand and gravel. Proc ICE. (78): 1325-1381 Chow, F.C. 1997. Investigations into displacement pile behaviour for offshore foundations. Ph.D Thesis, Imperial College London Connolly, T. 1998. Hollow Cylinder Tests on Dunquerque sand. Internal Report, Imperial College London

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Cuccovillo, T. and Coop, M.R. 1997. The measurement of local strains in triaxial testing using LVDTs. Géotechnique, 47 (1): 167-171. CUR 2001. Bearing capacity of steel pipe piles. Report 2001-8. Centre for Civil Engineering Research and Codes. Gouda, The Netherlands. Davies, P. 1975. Creep characteristics of three undisturbed clays. PhD Thesis, (Imperial College) University of London. DeJong, J.T., Randolph, M.F. & White, D.J. 2003. Interface load transfer degradation during cyclic loading: a microscale investigation Soils and Foundations, 43 (4). 91-94. Frank, R. 1994. Some recent developments on the behaviour of shallow foundations. General Report. 10th ECSMFE, Florence, Vol 4, Balkema: 1115-1146 Gasparre, A., Nishimura, S., Anh-Minh, N., Coop, M.R. & Jardine, R.J. 2007. The stiffness of natural London clay. Géotechnique, 57 (2): 33-48. Ho, Y.K., Jardine, R.J and Anh-Minh, N. 2011. Large displacement interface shear between steel and granular media. Géotechnique, 61 (3): 221-234. Hight, D.W., Gens A. and Symes, M.J. 1983. The development of a new hollow cylinder appparatus for investigating the effects of principal stress rotation in soils. Géotechnique, 33 (4): 355-384. Huang, A.B., and Hsu, H.H. 2005. Cone penetration tests under simulated field conditions. Géotechnique 55(5): 345–354. Ishihara, K., Tatsuoka, F. & Yasua, S. 1975. Undrained deformation and liquefaction of sand under cyclic stresses. Soils and Foundations, 15 (1): 29-44. Jardine, R.J. Symes, M.J.P.R. & Burland, J.B. 1984. The measurement of soil stiffness in the triaxial apparatus. Géotechnique 34 (3): 323-340. Jardine R. J., Potts D. M., Fourie A. B., and Burland J. B. 1986. Studies of the influence of non-linear stress-strain characteristics in soil-structure interaction. Géotechnique, 36, No 3, pp377-396. Jardine, R.J. and Potts, D.M. 1988. Hutton Tension Leg Platform foundations: an approach to the prediction of driven pile behaviour. Géotechnique, 38 (2): 231-252. Jardine, R.J. 1991. The cyclic behaviour of offshore piles. The Cyclic Loading of Soils, Eds. Brown & O'Reilly, Blackie & Son, Glasgow. Jardine, R.J. 1992. Observations on the kinematic nature of soil stiffness at small strains. Soils and Foundations, 32 (2): 111124. Jardine, R.J., Lehane, B.M. and Everton, S.J 1992. Friction coefficients for piles in sands and silts. Proc 3rd Int. Conf. on Offshore Site Investigations and Geotechnics, SUT London, Kluwer, Dordrecht, pp 661-677. Jardine, R.J. 1994. Offshore pile design for cyclic loading: North Sea clays. HSE Offshore Technology Report, OTN 94 157.85. Jardine R.J., Standing, J.R., Jardine, F.M., Bond, A.J. and Parker, E. 2001a. A competition to assess the reliability of pile prediction methods. Proc. XVth ICSMGE, Istanbul, Vol 2, pp 911-914 Jardine, R.J, Kuwano, R., Zdravkovic, L. and Thornton, C. 2001b. Some fundamental aspects of the pre-failure behaviour of granular soils. 2nd Int Symp. On Pre-failure Behaviour of Geomaterials, IS- Torino, Volume 2. Swets & Zeitlinger, Lisse, pp1077-1113. Jardine, R.J., Standing, J.R and Kovacevic, N. 2005a. Lessons learned from Full scale observations and the practical application of advanced testing and modelling. Proc International Symposium on Deformation Characteristics of Geomaterials, Lyon, Vol 2, Balkema, pp. 201-245. Jardine, R.J., Chow, FC, Overy, RF and Standing, J.R 2005b. ICP design methods for driven piles in sands and clays”. Thomas Telford, London p. 105.

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Jardine, R.J, Standing, J.R and Chow, F.C. 2006. Some observations of the effects of time on the capacity of piles driven in sand. Géotechnique 55 (4): 227-244. Jardine, R.J., Zhu, B., Foray, P. and Dalton, C.P. 2009. Experimental arrangements for the investigation of soil stresses developed around a displacement pile. Soils and Foundations; 49 (5): 661-673. Jardine, R.J., Andersen, K. and Puech, A. 2012. Cyclic loading of offshore piles: potential effects and practical design. Proc 7th Int. Conf. on Offshore Site Investigations and Geotechnics, SUT London, pp 59-100. Jardine R.J. and Standing, J.R. 2012. Field axial cyclic loading experiments on piles driven in sand. Soils and Foundations. 52 (4): 723-737. Jardine R.J, Zhu, B.T., Foray, P. and Yang, Z.X. 2013a. Measurement of Stresses around Closed-Ended Displacement Piles in Sand. Géotechnique 63 (1): 1–17. Jardine R.J, Zhu, B.T., Foray, P. and Yang, Z.X. 2013b. Interpretation of stress measurements made around closedended displacement piles in sand. Géotechnique, In Press. Kallehave, D., Le Blanc-Thilsted, C. and Liingard, M. 2012. Proc 7th Int. Conf. on Offshore Site Investigations and Geotechnics, SUT London, pp 465-472. Kovacevic, N. Jardine., R, Potts, D. Clukey, E. Brand, J.R. and Spikula, D. 2012. A numerical simulation of progressive slope failures generated by salt diaiprism combined with active sedimentation. Geotechnique. 62 (9): 777-786. Kuwano, R. and Jardine, R.J. 1998. Stiffness measurements in a stress path cell. Pre-failure behaviour of geomaterials. Thomas Telford, London, pp 391-395. Kuwano, R. 1999 The stiffness and yielding anisotropy of sand. PhD Thesis, Imperial College London Kuwano, R. and Jardine R.J. 2002a. On measuring creep behaviour in granular materials through triaxial testing. Canadian Geotechnical Journal; 39 (5): 1061-1074. Kuwano, R. and Jardine R.J. 2002b. On the applicability of cross anisotropic elasticity to granular materials at very small strains. Geotechnique, 52 (10): 727-750. Kuwano, R. and Jardine, R.J. 2007. A triaxial investigation of kinematic yielding in sand. Géotechnique, 57 (7): 563-580. Lehane, B.M., Jardine, R.J., Bond, A.J. and Frank, R. 1993. Mechanisms of shaft friction in sand from instrumented pile tests. ASCE Geot. Journal. 119 (1): 19-35. Lehane B.M., Schneider J.A. and Xu X. 2005. A review of design methods in offshore driven piles in siliceous sand. University of Western Australia (UWA) Report GEO 05358, 105p. Merritt, A., Schroeder, F., Jardine, R., Stuyts, B., Cathie, D., & Cleverly, W. 2012. Development of pile design methodology for an offshore wind farm in the North Sea. Proc 7th Int. Conf. on Offshore Site Investigations & Geotechnics, SUT, pp 439-448. Menkiti, C.O. 1995. Behaviour of clay and clayey-sand, with particular reference to principal stress rotation. PhD Thesis, University of London Muir-Wood, D. 2008. Critical states and soil modelling. Deformation Characteristics of Geomaterials. 1, IOS Amsterdam, 51-72 Nishimura, S. 2006. Laboratory study of the anisotropy of natural London Clay. PhD Thesis, Imperial College London. Nishimura, S., Minh, N.A. and Jardine, R.J. 2007. Shear strength anisotropy of natural London clay. Symposium in Print on Stiff Clays. Géotechnique, 57 (1), pp 49-62. Porovic, E. 1995. Investigations of soil behaviour using a resonant column torsional shear hollow cylinder apparatus. PhD Thesis, Imperial College London Potts, D. M. and Zdravkovic, L. 1999. Finite element analysis in geotechnical engineering: theory. Pub Thomas Telford, London, 440p.

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Potts, D. M. and Zdravkovic, L. 2001. Finite element analysis in geotechnical engineering: application. Pub Thomas Telford, London, 427p. Randolph, M. F. and Wroth, C. P. 1981. Application of the failure state in undrained simple shear to the shaft capacity in the driven piles. Géotechnique 31 (1): 143-157. Rimoy, S.P. and Jardine, R.J. 2011. Strain accumulation in a silica sand due to creep after normal compression, and during sustained low-level cyclic loading. Deformation Characteristics of Geomaterials. Proc. IS-Seoul, Hanrimwon, (1): 463-470. Rimoy, S.P., Jardine, R.J and Standing, J.R. 2013. Displacement response to axial cycling of piles driven in sand. Geotechnical Engineering, 116 (2): 131-146. Rimoy, S.P. 2013. Ageing and axial cyclic loading studies of displacement piles in sands. PhD Thesis, Imperial College London. Shen, C.K. 2013. A micromechanical investigation of drained simple shear tests on dense sand using Discrete Element Simulations. PhD Thesis, Imperial College London. Shibuya, S. & Hight, D.W. 1987. On the stress path in simple shear. Géotechnique 37 (4): 511–515. Shibuya, S., Hight, D.W. and Jardine, R.J. 2003a. Four Dimensional Local Boundary Surfaces of an Isotropically Consolidated Loose Sand. Soils and Foundations, 43 (2): 89103. Shibuya, S., Hight, D.W. and Jardine, R.J. 2003b. Local Boundary Surfaces of a loose sand dependent on consolidation path. Soils and Foundations 43 (3): 85-93. Sim, W.W., Aghakouchak, A. and Jardine, R.J. 2013. Effects of duration and amplitude on cyclic behaviour of overconsolidated sands under constant volume conditions. Geotechnical Engineering, 116 (2): 111-121. Takahashi, A. & Jardine, R.J. 2007. Assessment of standard research sand for laboratory testing, Quarterly Journal of Engineering Geology and Hydrogeology; 40 (1): 93-103. Tatsuoka, F., Jardine, R. J., Lo Presti, D., Di Benedetto, H. and Kodaka, T. 1999. Characterising the pre-failure deformation properties of geomaterials. Proc XIVth ICSMFE, Hamburg, Volume 4, Balkema, Vol 4 pp 2129-2164. Tatsuoka, F. 2011. Laboratory stress-strain tests for developments in geotechnical engineering. 1st Bishop Lecture, Deformation Characteristics of Geomaterials. Proc. IS-Seoul, Hanrimwon, Vol. 1, p 3-53. Terzahgi, K. and Peck, R.B. 1967. Soil mechanics in engineering practice. 2nd Ed.., New York, Wiley. Tsuha, C.H.C, Foray, P.Y., Jardine, R.J., Yang, Z.X., Silva, M. and Rimoy, S.P. 2012. Behaviour of displacement piles in sand under cyclic axial loading. Soils & Foundations, 52 (3): 393-410. Vucetic, M. 1994. Cyclic threshold shear strains in soils. Journal of Geotechnical Engineering, ASCE, 120 (12): 2208-2228. Yang, Z.X., Jardine, R.J., Zhu B.T., Foray, P. and Tsuha, C.H.C.. 2010. Sand grain crushing and interface shearing during displacement pile installation in sand, Géotechnique, 60 (6): 469-482. Yang, Z.X, Jardine, R.J, Zhu, B.T and Rimoy, S. 2013 The stresses developed round displacement piles penetrating in sand. Submitted to ASCE Geot. Journal. Zhang, C., Nguyen, G.D., & Einav, I. 2013. The end-bearing capacity of piles penetrating into crushable soils, Géotechnique, 63 (5): 341: 354. Zdravkovic L. and Jardine, R.J. 1997. Some anisotropic stiffness characteristics of a silt under general stress conditions. Géotechnique, 47 (3): 407-438. Zhu, B., Jardine, R.J. and Foray, P. 2009. The use of miniature soil stress measuring cells in laboratory applications involving stress reversals. Soils and Foundations; 49 (5): 675-688.

Ishihara Lecture Ishihara Lecture Soil-Foundation-Structure Systems beyond Conventional Seismic Failure Soil-Foundation-Structure Systems Beyond Conventional Seismic Failure Thresholds Thresholds Conférence Ishihara Conférence Ishihara Les systèmes sol-fondation-structure qui dépassent les limites de la rupture parasismique Les systèmes conventionnellesol-fondation-structure qui dépassent les limites de la rupture parasismique conventionnelle Gazetas G. Gazetas Professor,G. National Technical University of Athens, Greece Professor, National Technical University of Athens, Greece

ABSTRACT: A new paradigm has now emerged in performance–based seismic design of soilfoundationstructure systems. Instead of imposing strict safety limits on forces and moments transmitted from the foundation onto the soil (aiming at avoiding pseudo-static failure), the new dynamic approach “invites” the creation of two simultaneous “failure” mechanisms: substantial foundation uplifting and ultimate-bearing-capacity slippage, while ensuring that peak and residual deformations are acceptable. The paper shows that allowing the foundation to work at such extreme conditions not only may not lead to system collapse, but it would help protect (save) the structure from seismic damage. A potential price to pay: residual settlement and rotation, which could be abated with a number of foundation and soil improvements. Numerical studies and experiments demonstrate that the consequences of such daring foundation design would likely be quite beneficial to bridge piers and building frames. It is shown that system collapse could be avoided even under seismic shaking far beyond the design ground motion. RÉSUMÉ : Un nouveau paradigme a émergé dans la conception sismique de la performance des systèmes sol – fondation – structure. Au lieu d'imposer des coefficients de sûreté sur les forces et les moments transmis par la fondation sur le sol (pour éviter la rupture pseudo-statique), la nouvelle approche dynamique permet la création de deux modes de rupture simultanés : le soulèvement important de la fondation et le dépassement de la capacité portante ultime, tout en assurant que les déformations maximales et résiduelles sont acceptables. L’article montre que, quand on permet à la fondation de travailler dans ces conditions extrêmes, l'effondrement du système peut être évité et de plus la structure peut être protégée du dommage sismique. Un prix potentiel à payer : le déplacement et la rotation résiduels, qui peuvent être contrôlés avec différentes méthodes d'amélioration de la fondation et des sols. Des études numériques et expérimentales montrent que les conséquences d'une telle conception audacieuse de la fondation seraient certainement très bénéfiques pour les ponts et les bâtiments. On montre que l'effondrement du système pourrait être évité, même pendant des secousses sismiques qui dépassent le mouvement de calcul. KEYWORDS: seismic analysis, performance-based design, foundation rocking, bearing capacity failure, nonlinear vibrations 1

CURRENT STATE OF PRACTICE: THE CONVENTIONAL “WISDOM”

 sliding at the soil–footing interface or excessive uplifting of a shallow foundation  passive failure along the normal compressing sides of an embedded foundation  a combination of two or more of the above “failure” modes. In this conventional approach to foundation design, “overstrength” factors plus (explicit and implicit) factors of safety larger than 1 (e.g. in the form of “material” factors) are introduced against each of the above “failure” modes, in a way qualitatively similar to the factors of safety of the traditional static design. Thus, the engineer is certain that foundation performance will be satisfactory and there will be no need to inspect and repair after strong earthquake shaking  a task practically considered next to impossible. Some of the above thresholds stem not just from an understandable engineering conservatism, but also from a purely (pseudo) static thinking. It will be shown that such an approach may lead not only to unnecessarily expensive foundation solutions but also, in many situations, to less safe structures.

Seismic design of structures recognises that highly inelastic material response is unavoidable under the strongest possible shaking of the particular location and for the specific soil where the structure is founded. “Ductility” levels of the order of 3 or more are usually allowed to develop under seismic loading, implying that the strength of a number of critical bearing elements is fully mobilized. In the prevailing structural terminology “plastic hinging” is allowed to develop as long as the overall stability is maintained. By contrast, a crucial goal of current practice in seismic “foundation” design, particularly as entrenched in the respective codes is to avoid the mobilisation of “strength” in the foundation. In the words of EC8 (Part 2, § 5.8) : “…foundations shall not be used as sources of hysteretic energy dissipation, and therefore shall be designed to remain elastic under the design seismic action.” In structural terminology : no “plastic hinging” is allowed in the foundation. In simple geotechnical terms, the designer must ensure that the below-ground (and hence un-inspectable) support system will not even reach a number of “thresholds” that would conventionally imply failure. Specifically, the following states are prohibited :  plastic structural “hinging” in piles, pile-caps, foundation beams, rafts, and so on  mobilisation of the so-called bearing-capacity failure mechanisms under cyclicallyuplifting shallow foundations

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2 SOME COMPELLING REASONS TO GO BEYOND CONVENTIONAL THRESHOLDS

ignored, even if their geometrically–nonlinear nature presents computational difficulties. In fact, it is worthy of note that the lack of recognition of the fundamental difference between pseudo-static and seismic overturning threshold accelerations has led humanity to a gross under-estimation of the largest ground accelerations that must have taken place in historic destructive earthquakes. Because, by observing in numerous earthquakes that very slender blocks (of width b and height h, with h >> b) or monuments in precarious equilibrium that had not overturned, engineers had invariably attributed the fact to very small peak accelerations, less than (b/h)g, as would be necessary if accelerations were applied pseudostatically in one direction. Today we know that sometimes even five times as large peak ground acceleration of a high-frequency motion may not be enough to overturn a slender block (Koh et al 1986, Makris & Roussos 2000, Gazetas 2001). Simply stated: even severe uplifting (conventional “failure”) may not lead to overturning (true “collapse”) under dynamic seismic base excitation. (d) Compatibility with structural design is another reason for the soil−structure interaction analyst to compute the lateral load needed for collapse of the foundation system, as well as (in more detail) the complete load–displacement or moment– rotation response to progressively increasing loading up to collapse. Indeed, in State of the Art (SOA) structural engineering use is made of the so-called “pushover” analysis, which in order to be complete requires the development of such information from the foundation analyst. In addition to the above “theoretical” arguments, there is a growing need for estimating the “collapse motion” : insurance coverage of major construction facilities is sometimes based on estimated losses under the worst possible (as opposed to probable) earthquake scenario. (e) Several persuasive arguments could be advanced on the need not to disallow structural plastic “hinging” of piles: • Yielding and cracking of piles (at various critical depths) is unavoidable with strong seismic shaking in soft soils, as the Kobe 1995 earthquake has amply revealed. • Refuting the contrary universal belief, post-earthquake inspection of piles is often feasible (with internally placed inclinometers, borehole cameras, integrity shock testing, under-excavation with visual inspection ), although certainly not a trivial operation. Again, Kobe offered numerous examples to this effect. • The lateral confinement provided by the soil plays a very significant role in pile response, by retarding the development of high levels of localised plastic rotation, thereby providing an increase in ductility capacity. Sufficient displacement ductility may be achieved in a pile shaft with transverse reinforcement ratio as low as 0.003 (Butek et al 2004). • The presence of soil confinement leads to increased plastic hinge lengths, thus preventing high localised curvatures (Tassios 1998). Therefore, the piles retain much of their axial load carrying capacity after yielding. Thus, a broadly distributed plastic deformation on the pile may reduce the concentrated plastification on the structural column  so detrimental to safety. Furthermore, when subjected to strong cyclic overturning moment, end-bearing piles in tension will easily reach their full frictional uplifting capacity. It has been shown analytically and experimentally that this does not imply failure. The same argument applies to deeply embedded (caisson) foundations. (f) The current trend in structural earthquake engineering calls for a philosophical change : from strength-based design (involving force considerations) to performance-based design (involving displacement considerations) [Pauley 2002, Priestley et al 2000, 2003, Calvi 2007]. Geotechnical earthquake engineering has also been slowly moving towards performance–based seismic design: gravity retaining structures

A growing body of evidence suggests that soil–foundation plastic yielding under seismic excitation is unavoidable, and at times even desirable; hence, it must be considered in analysis and perhaps allowed in design. [See for an early recognition : Pecker 1998, Faccioli & Paolucci 1999, Martin & Lam 2000, FEMA-356 2000, Kutter et al 2001, Gazetas & Apostolou 2003.] The urgent need to explicitly consider the possibility of the foundation system to go beyond “failure” thresholds, and the potential usefulness of doing so, have emerged from : (a) The large (often huge) effective ground acceleration, A, and velocity, V, levels recorded in several earthquakes in the last 25 years. A few examples : • 1994 Ms ≈ 6.8 Northridge : A = 0.98 g, V = 140 cm/s ; • 1995 MJ ≈ 7.2 Kobe : A = 0.85 g, V = 120 cm/s ; • 1986 Ms ≈ 5.6 San Salvador : A = 0.75 g, V = 84 cm/s ; • 2003 Ms = 6.4 Lefkada : A ≈ 0.55 g, V = 50 cm/s ; • 2007 MJ ≈ 6.9 Niigata : A =1.20 g, V = 100cm/s . With the correspondingly large accelerations in the (above– ground) structure from such ground motions (spectral Sa values well in excess of 1 g), preventing “plastic hinging” in the foundation system is a formidable task. And in fact, it may not even be desirable: enormous ductility demands might be imposed to the structure if soil–foundation “yielding” would not take place to effectively limit the transmitted accelerations. Several present-day critically–important structures on relatively loose soil could not have survived severe ground shaking if “plastic hinging” of some sort had not taken place in the “foundation”  usually unintentionally. (b) In seismically retrofitting a building or a bridge, allowing for soil and foundation yielding is often the most rational alternative. Because increasing the structural capacity of some elements, or introducing some new stiff elements, would then imply that the forces transmitted onto their foundation will be increased, to the point that it might not be technically or economically feasible to undertake them “elastically”. The new American retrofit design guidelines (FEMA 356) explicitly permit some forms of inelastic deformations in the foundation. A simple hypothetical example referring to an existing three– bay multi–story building frame which is to be retrofitted with a single–bay concrete “shear” wall had been introduced by Martin & Lam 2000. Such a wall, being much stiffer than the columns of the frame, would carry most of the inertia-driven shear force and would thus transmit a disproportionately large horizontal force and overturning moment onto the foundation compared with its respective small vertical force. If uplifting, sliding, and mobilisation of bearing capacity failure mechanisms in the foundation had been all spuriously ignored, or had been conversely correctly taken into account, would have led to dramatically different results. With “beyond–threshold” action in the foundation the shear wall would “shed” off some of the load onto the columns of the frame, which must then be properly reinforced ; the opposite would be true when such action (beyond the thresholds) is disallowed. The Engineer therefore should be able to compute the consequences of “plastic hinging” in the foundation before deciding whether such “hinging” must be accepted, modified, or avoided (through foundation changes). (c) Many slender historical monuments (e.g. ancient columns, towers, sculptures) may have survived strong seismic shaking during their life (often of thousands of years). While under static conditions such “structures” would have easily toppled, it appears that sliding at, and especially uplifting from, their base during oscillatory seismic motion was a key to their survival (Makris & Roussos 2000, Papantonopoulos 2000). These nonlinear interface phenomena cannot therefore be

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Honour Lectures / Conférences honorifiques

are indeed allowed to slide during the design earthquake. The time is therefore ripe for soil–foundation–structure interaction (SFSI) to also move from imposing “safe” limits on forces and moments acting on the foundation (aiming at avoiding pseudostatic “failure”) to performance–based design in which all possible conventional “failure” mechanisms are allowed to develop, to the extent that maximum and permanent displacements and rotations are kept within acceptable limits.

4 ROTATIONAL MONOTONIC RESPONSE OF SHALLOW FOUNDATIONS Much of the research in earlier years on dynamic rocking of foundations and dynamic soil−structure interaction had focused on linear response. Elastic stiffness and damping as functions of frequency have been developed and utilised to describe the dynamic action of the foundation system. The various US seismic codes in the last 30+ years have promulgated linear approximations to deal with seismic soil−structure interaction. The behavior of “Rocking Foundations” significantly deviates from linear visco-elasticity: uplifting introduces strong geometric nonlinearity and even damping due to impact ; soil yielding and plastic deformation generate hysteresis, implying significant frequency-independent damping, while when bearing-capacity slippage mechanisms develop a limiting plateau restricts the passage of high accelerations from the ground into the superstructure. In monotonic loading, a most crucial parameter controlling the moment−rotation, M−θ, relation of a specific foundation is the factor of safety against vertical static bearing capacity failure :

3 THE CONCEPT OF “ROCKING ISOLATION” IN FOUNDATION DESIGN The paper addresses the case of structure-foundation systems oscillating mainly in a rotational mode (rocking). Subjected to strong seismic shaking, structures tend to experience large inertial forces. For tall-slender structures these forces will lead to overturning moments onto the foundation that may be disproportionally large compared to the vertical load. As a result, a shallow foundation may experience detachment (uplifting) of one edge from the supporting soil. This in turn will lead to increased normal stresses under the opposite edge of the foundation. Development of a bearing capacity failure mechanism is quite possible if such a concentration leads to sufficiently large stresses. But, in contrast to a static situation, even then failure may not occur. Thanks to the cyclic and kinematic nature of earthquake induced vibrations : (i) the inertial forces do not act “forever” in the same direction to cause failure (as would be the case with static load), but being cyclic, very soon reverse and thereby relieve the distressed soil; and (ii) the developing inertial forces are not externally applied predetermined loads, but are themselves reduced once the soil-foundation system reaches its (limited) ultimate resistance  the foundation system acts like a fuse. As a result, the system experiences nonlinear-inelastic rocking oscillations, which may or may not result in excessive settlement and rotation. But failure is almost unlikely. In the last 10 years a number of research efforts have explored the consequences of substantial foundation rocking on the response of the supported structure, theoretically and experimentally : Kutter et al 2003, Gajan et al 2005, Harden et al 2006, Kawashima et al 2007, Apostolou et al 2007, Paolucci et al 2008, Chatzigogos & Pecker 2010, Deng et al 2012. The results of these studies confirmed the idea that stronglynonlinear rocking oscillations under seismic excitation can be of benefit to the structure. Taking the whole idea one small step farther, it is proposed that the design of a shallow foundation should actively “invite” the creation of two simultaneous “failure” mechanisms: substantial foundation uplifting and ultimate bearing-capacity sliding. This would be accomplished by substantially underdesigning the foundation  e.g., by reducing its width and length to, say, one-half of the values required with current design criteria. This can be thought of as a reversal of the “capacity” design: “plastic hinging” will take place in the foundation-soil system and not at the column(s) of the structure. Fig. 1 elucidates the main idea of Rocking Isolation. The benefits of designing the foundation to work at and beyond its conventional limits will become evident in the sequel. To this end, three examples will elucidate the dynamics of “Rocking Isolation” in comparison with the dynamics of the conventional design :

Fs = Nuo/N

(1)

where Nuo is the ultimate load under purely vertical loading and N the acting vertical load. Fig. 2 offers typical results for a homogeneous (G and su ) soil for three Fs values : a very high one (20), a low one (2), and an extremely low one (1.25). M is normalized by Nuo B, where B is the width of the footing in the direction of loading. This leads to curves which, for the homogeneous profile considered, depend solely on the so-called “rigidity index”, G/ su , and the shape of the footing. Also shown in Fig. 2 are the snapshots of the deformed soil and the contours of plastic strain as they develop when the maximum moment is reached  apparently at different angles of rotation. The following are worthy of note in the figure: • The foundation with Fs = 20 (which can be interpreted either as a very-lightly loaded foundation or as a “normally”-loaded foundation on very stiff soil) despite its largest initial elastic rocking stiffness fails at the smallest value of applied moment: Mu ≈ 0.025 Nuo B

(2a)

Indeed if Fs → ∞ , i.e. there is no vertical load onto the foundation, Mu would vanish, due to the tensionless nature of the soil−footing interface. • As expected from the literature (Meyerhof 1963, Georgiadis and Butterfield 1988, Salençon and Pecker 1995, Αllotey and Naggar 2003, Apostolou and Gazetas 2005, Gajan and Kutter 2008, Chatzigogos et al. 2009, Gouvernec 2009, Gajan and Kutter 2008) the largest maximum moment is attained by the Fs = 2 footing : Mu ≈ 0.13 Nuo B

(2b)

but its elastic initial rocking stiffness is smaller than for the Fs = 20 foundation. Evidently, the extensive plastic deformations upon the application of the vertical (heavy) load soften the soil so that a small applied moment meets less resistance  hence lower stiffness. However, Fs = 2 achieves the largest ultimate Mu as it leads to an optimum combination of uplifting and bearing-capacity mobilization.

(a) a bridge pier, free to rotate at its top (b) a two-storey two-bay asymmetric frame (MRF) (c) a three-storey retrofitted frame−shearwall structure.

• A more severely loaded foundation, however, with the (rather unrealistic) Fs = 1.25 will only enjoy an even smaller initial stiffness and a smaller ultimate moment than the Fs = 2 foundation. Notice that in this case no uplifting accompanies the plasticification of the soil.

In each case, the two alternatives ( the conventional and the rocking-isolated system) are subjected to numerous acceleration time histories the overall intensity of which is either within or well beyond the design earthquake levels.

The failure envelope (also called interaction diagram) in NM space is given in Fig. 3 for the specific example. It was

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obtained with the same numerical (FE) analysis as the curves and snapshots of Fig. 2, and can be expressed analytically as a function of the static factor of safety (FS) as

 =

  1 −     

The following relationship has been developed from FE results by Kourkoulis et al, 2012, for the overturning angle θc = θc(Fs) :

(3)

The specific plot is in terms of N/Nuo which is 1/Fs which ranges between 0 and 1. Notice that heavily and lightly loaded foundations with 1/Fs symmetrically located about the 1/FS = 0.5 value where the Mu is the largest, have the same moment capacity : yet their behavior especially in cyclic loading is quite different as will be shown subsequently.

6

An increasingly popular concept in structural earthquake engineering is the so-called “pushover” analysis. It refers to the nonlinear lateral force-displacement relationship of a particular structure subjected to monotonically increasing loading up to failure. The development (theoretical or experimental) of such pushover relationships has served as a key in simplified dynamic response analyses that estimate seismic deformation demands and their ultimate capacity. We apply the pushover idea to a shallow foundation supporting an elevated mass, which represents a tall slender structure with h/B = 2 (or “slenderness” ratio h/b = 4, where b = B/2). This mass is subjected to a progressively increasing horizontal displacement until failure by overturning. Since our interest at this stage is only in the behavior of the foundation, the structural column is considered absolutely rigid. The results are shown in Fig:4(a) and (b) for two Fs values : 5 and 2. The difference in the M-θ response curves from those of Fig. 2 stems from the so-called P-δ effect. As the induced lateral displacement of the mass becomes substantial its weight induces an additional aggravating moment, mgu = mgθh, where θ is the angle of foundation rotation. Whereas before the ultimate moment Mu is reached the angles of rotation are small and this aggravation is negligible, its role becomes increasingly significant at larger rotation and eventually becomes crucial in driving the system to collapse. Thus, the (rotation controlled) M-θ curve decreases with θ until the system topples at an angle θc . This critical angle for a rigid structure on a rigid base (FS = ∞) is simply :  

(4)

where b = the foundation halfwidth. For very slender systems the approximation

,∞ ≈

 



  





(5)

CYCLIC RESPONSE ACCOUNTING FOR P−δ EFFECTS

Slow cyclic analytical results are shown for the two aforementioned systems having static factors of safety (FS = 5 and 2). The displacement imposed on the mass center increased gradually; the last cycle persisted until about 4 or 5 times the angle θu of the maximum resisting moment. As can be seen in the moment−rotation diagrams, the loops of the cyclic analyses for the safety factor FS = 5 are well enveloped by the monotonic pushover curves in Figure 7(a). In fact, the monotonic and maximum cyclic curves are indistinguishable. This can be explained by the fact that the plastic deformations that take place under the edges of the foundation during the deformationcontrolled cyclic loading are too small to affect to any appreciable degree of response of the system when the deformation alters direction. As a consequence, the residual rotation almost vanishes after a complete set of cycles ― an important (and desirable) characteristic. The system largely rebounds, helped by the restoring role of the weight. A key factor of such behaviour is the rather small extent of soil plastification, thanks to the light vertical load on the foundation. The cyclic response for the FS = 2 system is also essentially enveloped by the monotonic pushover curves. However, there appears to be a slight overstrength of the cyclic “envelope” above the monotonic curve. For an explanation see Panagiotidou et al, 2012. But the largest difference between monotonic and cyclic, on one hand, and FS = 2 and 5, on the other, is in the developing settlement. Indeed, monotonic loading leads to monotonicallyupward movement (“heave”) of the center of the FS = 5 foundation, and slight monotonically-downward movement (“settlement”) of the FS = 2 foundation. Cyclic loading with FS = 5 produces vertical movement of the footing which follows closely its monotonic upheaval. But the FS = 5 foundation experiences a progressively accumulating settlement  much larger that its monotonic settlement would have hinted at. The hysteresis loops are now wider. Residual rotation may appear upon a full cycle of loading, as inelastic deformations in the soil are now substantial. The above behavior is qualitatively similar to the results of centrifuge experiments conducted at the University of California at Davis on sand and clay (e.g., Kutter et al. 2003, Gajan et al. 2005) large-scale tests conducted at the European Joint Research Centre, (Negro et al. 2000, Faccioli et al. 1998), and 1-g Shaking Table tests in our laboratory at the National Technical University of Athens on sand (Anastasopoulos et al 2011, 2013, Drosos et al 2012). In conclusion, the cyclic moment−rotation behavior of foundations on clay and sand exhibits to varying degrees three important characteristics with increasing number of cycles : • no “strength” degradation (experimentally verified). • sufficient energy dissipation  large for small FS values, smaller but still appreciable for large ones. (Loss of energy due to impact will further enhance damping in the latter category, when dynamic response comes into play.) • relatively low residual drift especially for large FS values  implying a re-centering capability of the rocking foundation. These positive attributes not only help in explaining the favorable behavior of “Rocking Foundation”, but also enhance the reliability of the geotechnical design.

5 MONOTONIC RESPONSE ACCOUNTING FOR P−δ EFFECTS

,∞ = 

   ≈ 1 −  + 1 − ,∞  

(4a)

is worth remembering. As the static vertical safety factor (FS) diminishes, the rotation angle (θc) at the state of imminent collapse (“critical” overturning rotation) also slowly decreases. Indeed, for rocking on compliant soil, θc is always lower than it is on a rigid base (given with Eq. 4). For stiff elastic soil (or with a very large static vertical safety factor) θc is imperceptibly smaller than that given by Eq. 4, because the soil deforms slightly, only below the (right) edge of the footing, and hence only insignificantly alters the geometry of the system at the point of overturning. As the soil becomes softer, soil inelasticity starts playing a role in further reducing θc. However, such a reduction is small as long as the factor of safety (FS) remains high (say, in excess of 3). Such behaviour changes drastically with a very small FS: then the soil responds in strongly inelastic fashion, a symmetric bearing-capacity failure mechanism under the vertical load N is almost fully developed, replacing uplifting as the prevailing mechanism leading to collapse θc tends to zero.

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Honour Lectures / Conférences honorifiques

7 SEISMIC RESPONSE OF BRIDGE PIER ON SHALLOW FOUNDATION

failure mechanisms in the underlying soil, leaving the superstructure totally intact. Notice that the red regions of large plastic shearing are of great extent, covering both half-widths of the foundation and indicating alternating mobilization of the bearing capacity failure mechanisms, left and right. The above observations are further confirmed by the time history of deck drift shown in Fig. 5(c). The two components of drift, are shown, one due to footing rotation in blue and one due to structural distortion in green. Their sum is shown in red. Evidently, the conventional design experiences essentially only structural distortion which leads to uncontrollable drifting  collapse. In marked contrast, the system designed according to the new philosophy easily survives. It experiences substantial maximum deck drift (about 40 cm), almost exclusively due to foundation rotation. Nevertheless, the residual foundation rotation leads to a tolerable 7 cm deck horizontal displacement at the end of shaking. Fig. 5(d) further elucidates the action of the foundation-soil system. The M-θ relationship shows for the 11m2 foundation a nearly linear viscoelastic response, well below its ultimate capacity and apparently with no uplifting. On the contrary, the 7m2 (under-designed) foundation responds well past its ultimate moment capacity, reaching a maximum θ ≈ 30 mrad, generating hysteretic energy dissipation, but returning almost to its original position, i.e. with a negligible residual rotation. However, energy dissipation is attained at a cost : increased foundation settlement. While the practically elastic response of the conventional (over-designed) foundation leads to a minor 4 cm settlement, the under-designed foundation experiences an increased accumulated 15 cm settlement. Although such settlement is certainly not negligible, it can be considered as a small price to pay to avoid collapse under such a severe ground shaking. Perhaps not entirely fortuitously, the residual rotation in this particular case turned out to be insignificant. The recentering capability of the design certainly played some role in it.

The concept of “Rocking Isolation” is illustrated in Fig. 5 by comparing the response of a 12 m tall bridge pier carrying a deck of four lanes of traffic for a span of about 35 m  typical of elevated highways around the world. The bridge chosen for analysis is similar to the Hanshin Expressway Fukae bridge, which collapsed spectacularly in the Kobe 1995 earthquake. The example bridge is designed in accordance to (EC8 2000) for a design acceleration A = 0.30 g, considering a (ductility-based) behavior factor q = 2. With an elastic (fixed-base) vibration period T = 0.48 sec the resulting design bending moment MCOL ≈ 45 MNm. The pier is founded through a square foundation of width B on an idealized homogeneous 25 m deep stiff clay layer, of undrained shear strength su = 150 kPa (representative soil conditions for which a surface foundation would be a realistic solution). Two different foundation widths are considered to represent the two alternative design approaches. A large square foundation, B = 11 m, is designed in compliance with conventional capacity design, applying an overstrength factor γRd = 1.4 to ensure that the plastic “hinge” will develop in the superstructure (base of pier). Taking account of maximum allowable uplift (eccentricity e = M / N < B/3, where N is the vertical load), the resulting safety factors for static and seismic loading are FS = 5.6 and FE = 2.0, respectively. A smaller, under-designed, B = 7 m foundation is considered in the spirit of the new design philosophy. Its static safety factor FS= 2.8, but it is designed applying an “understrength” factor 1/1.4 ≈ 0.7 for seismic loading. Thus, the resulting safety factor for seismic loading is lower than 1.0 (FE ≈ 0.7). The seismic performance of the two alternatives is investigated through nonlinear FE dynamic time history analysis. An ensemble of 29 real accelerograms is used as seismic excitation of the soil–foundation–structure system. In all cases, the seismic excitation is applied at the bedrock level. Details about the numerical models and the requisite constitutive relations can be seen in Anastasopoulos et al, 2010, 2011. Results are shown here only for a severe seismic shaking, exceeding the design limits: the Takatori accelerogram of the 1995 MJMA 7.2 Kobe earthquake. With a direct economic loss of more than $100 billion, the Kobe earthquake needs no introduction. Constituting the greatest earthquake disaster in Japan since the 1923 Ms = 8 Kanto earthquake, it is simply considered as one of the most devastating earthquakes of modern times. Of special interest is the damage inflicted to the bridges of Hanshin Expressway, which ranged from collapse to severe damage. The aforementioned bridge chosen for our analysis is very similar to the Fukae section of Hanshin Expressway, 630 m of which collapsed during the earthquake of 1995. It is therefore logical to consider this as a reasonably realistic example of an “above the limits” earthquake. In particular, the Takatori record constitutes one of the worst seismic motions ever recorded : PGA = 0.70 g, PGV = 169 cm/s, bearing the “mark” of forward rupture directivity and of soil amplification. Fig. 5 compares the response of the two alternatives, in terms of deformed mesh at the end of shaking with superimposed the plastic strains. In the conventionally designed system there is very little inelastic action in the soil; the red regions of large plastic deformation are seen only under the severely “battered” edges of the rocking foundation  but without extending below the foundation. “Plastic hinging” forms at the base of the pier, leading to a rather intense accumulation of curvature (deformation scale factor = 2).The P−δ effect of the mass will further aggravate the plastic deformation of the column, leading to collapse. In stark contrast, with the new design scheme the “plastic hinge” takes the form of mobilization of the bearing capacity

8 SEISMIC RESPONSE OF TWO−STOREY TWO BAY ASYMMETRIC FRAME The frame of Fig. 6 was structural designed according to EC8 for an effective ground acceleration A = 0.36 g and ductilitydependent “behavior” factor q = 3.9. The soil remains the stiff clay of the previous example. Two alternative foundation schemes are shown in the figure . The conventionally over-designed footings can mobilize a maximum moment resistance Mu from the underlying soil, larger than the bending moment capacity of the corresponding column MCOL .. For static vertical loads, a factor of safety FS ≥ 3 is required against bearing capacity failure. For seismic load combinations, a factor of safety FE = 1 is acceptable. In the latter case, a maximum allowable eccentricity criterion is also enforced: e = M/N ≤ B/3. For the investigated soil–structure system this eccentricity criterion was found to be the controlling one, leading to minimum required footing widths B = 2.7 m, 2.5 m and 2.4 m for the left, middle, and right footing, respectively. Bearing capacities and safety factors are computed according to the provisions of EC8, which are basically similar to those typically used in foundation design practice around the world. The under-sized footings of the rocking isolation scheme, are “weaker” than the superstructure, guiding the plastic hinge to or below the soil–footing interface, instead of at the base of the columns. The small width of the footings promotes full mobilization of foundation moment capacity with substantial uplifting. The eccentricity criterion is completely relaxed, while FE < 1 is allowed. The static FS ≥ 3 remains a requirement as a measure against uncertainties regarding soil strength. Moreover, it turns out that FS ≥ 4 might be desirable in order to promote uplifting–dominated response, and thereby limit seismic settlements [Kutter et al. 2003, Faccioli et al. 2001,Pecker &

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Pender 2000, Kawashima et al. 2007, Chatzigogos et al. 2009; Panagiotidou et al. 2012]. Applying the methodology which has been outlined in Gelagoti et al. 2012, the footings were designed to be adequately small to promote uplifting, but large enough to limit the settlements. Aiming to minimize differential settlements stemming from asymmetry, the three footings were dimensioned in such a manner so as to have the same FS. Based on the above criteria, the resulting footing widths for the rocking–isolated design alternative are B = 1.1 m, 1.8 m, and 1.3 m, for the left, middle, and right footing, respectively: indeed, substantially smaller than those of the code-based design. Footing dimensions and static factors of safety against vertical loading of the two designs are summarized in Table 1.

(≈ 3 cm/1.2 m) for the two side footings and 0.033 (≈ 6 cm/1.8 m) for the central one, the latter is substantially larger in width and hence its settlement is larger in absolute terms. Naturally, the three footings are not subjected to exactly the same loading, something which further complicates the response. Such differential settlements may inflict additional distress in the superstructure, and are therefore worthy of further investigation. 9 THREE−STOREY FRAME RETROFITTED WITH SHEAR−WALL

Table 1. Footing dimensions and corresponding factors of safety (computed following the provisions of EC8) against vertical loading for the seismic load combination (G + 0.3Q) for the two design alternatives of Fig. 6. Conventional Design

Rocking Isolation

Footing

B (m)

FS

Footing

B (m)

FS

Left

2.7

32.6

Left

1.1

5.4

Middle

2.5

10.6

Middle

1.8

5.4

Right

2.4

18.1

Right

1.3

5.4

The performance of the two design alternatives is compared in Fig. 6. The deformed mesh with superimposed plastic strain contours of the two alternatives is portrayed on top (Fig. 6a). With the relentless seismic shaking of the Takatori motion, the conventionally designed frame collapses under its gravity load (due to excessive drift of the structure, the moments produced by P–δ effects cannot be sustained by the columns, leading to loss of stability and total collapse). As expected, plastic hinges firstly develop in the beams and subsequently at the base of the three columns, while soil under the footings remains practically elastic. The collapse is also evidenced by the substantial exceedance of the available curvature ductility of the columns (Fig. 6b). Conversely, the rocking–isolated frame withstands the shaking, with plastic hinging taking place only in the beams, leaving the columns almost unscathed (moment-curvature response: elastic). Instead, plastic hinging now develops within the underlying soil in the form of extended soil plastification (indicated by the red regions under the foundation. The time histories of inter-storey drift further elucidate the aforementioned behavior of the two design alternatives (Fig. 6d). Thanks to the larger bending moment capacity of the column than of the footing, damage is guided “below ground” and at the soil–foundation interface in the form of detachment and uplifting  evidenced in Fig. 6d by the zero residual rotation, unveiling the re-centering capability of the under-designed foundation scheme. The price to pay: large accumulated settlements. Moreover, despite the fact that the three footings have been dimensioned to have the same static factor of safety FS (in an attempt to minimize differential settlements exacerbated from asymmetry), the central footing settles more than the two side footings, leading to a differential settlement of the order of 3 cm. The difference in the settlement stems of course from their differences in width. As previously discussed, the central footing was made larger (B = 1.8 m, compared to 1.1 m and 1.3 m of the two side footings) in order to maintain the same FS. Since the latter is common for the three footings, if the loading is more-or-less the same, their response should be similar. However, such equivalence refers to dimensionless quantities, not absolute values [see Kourkoulis et al., 2012b]. In other words, while the three footings sustain almost the same dimensionless settlement w/B, which is roughly equal to 0.025

60

The results presented now are not from numerical analysis as the previous one, but from Shaking Table experiments. They refer to a 3-storey two-bay frame which was designed according to the pre-1970 seismic regulations, for a base shear coefficient of 0.06. Because of the small value of this coefficient and the otherwise inadequate design, the frame has columns of crosssection 25 x 25 cm2 and beams 25 x 50 cm2 resulting in a strong beam−weak column system. Naturally, it fails by first “softstory” type of collapse when excited by motions corresponding to today’s codes with effective ground accelerations of the order of 0.30g and more. To upgrade the frame, a strong and stiff Shear Wall 1.5 m x 0.3 m in cross-section is constructed replacing the middle column, as shown in Fig. 7. The 1:10−scale model is supported on dense fine−grained Dr ≈ 80% sand. The original footings of all three columns were 1.5 m square. For the retrofitted frame the two columns retained their original 1.5 x 1.5m2 footings. The foundation of the Shear Wall (SW) is of special geotechnical interest : due to its disproportionately large lateral stiffness the SW tends to attract most of the seismically induced shear force and hence to transmit onto the foundation a large overturning moment. By contrast, its vertical load is relatively small. To meet the eccentricity limit e = M/N < B/3, a large foundation 6.0m x 0.80 m is thus necessary. Hence, the conventional solution of Fig. 8. Of course the resulting vertical bearing-capacity factor of safety is unavoidably large, FS ≅ 10, and the seismic apparent factor of safety against moment bearing-capacity is also far more than adequate : FE = 2. The decision to reduce the footing width to merely B = 3.5 m is not only economically favorable, but in the harsh reality of old buildings it may often be the only feasible decision in view of the usual space limitations due to pipes, small basements, walls, etc, present in the base. We will see if it is also favorable technically in resisting a strong seismic shaking. To be practical, in the above sense, no change is made to the column footings. (1.5 m square). We subject all three structures [ i.e., “a” the original frame, “b” the retrofitted with a SW founded on conventionallyconservative footing, and “c” the retrofitted with the underdesigned SW footing] to a number of strong ground excitations. Frame “a” easily fails as sketched in Fig. 8, where the physical collapse was artificially prevented by an external protective barrier in the Shaking Table experiment. The conventionally retrofitted SW-frame “b” could withstand most excitations. But with some of the strongest motions it developed substantial plastification at its base and led to residual top drift of an unacceptable 8%. The unconventionally–founded system “c” behaved much better with residual top drift of merely 2%. Figure 8 sketches the deformation pattern of the three systems while Fig. 7 plots the time histories of structural−distortion and foundation−rotation induced top drift ratio. It is seen that not only is the total drift of the RockingIsolated system only 2% but at least half of it is solely due to foundation rotation, rather than damage to the SW. The penalty to pay is the increased settlement (1.5 cm rather 0.8 cm) which nevertheless in this particular case would be acceptable for most applications.

Honour Lectures / Conférences honorifiques

10 CONCLUSIONS

12 REFERENCES

(a) Current seismic design practice leads most often to very conservative foundation solutions. Not only are such foundations un-economical but are sometimes difficult to implement. Most significantly : they are agents of transmitting large accelerations up to the superstructure. The ensuing large inertial forces send back in “return” large overturning moments (and shear forces) onto the foundation  a vicious circle.

Allotey N., El Naggar M.H. 2003. Analytical moment–rotation curves for rigid foundations based on a Winkler model. Soil Dynamics and Earthquake Engineering, 23, 367–381. Allotey N., El Naggar M.H. 2007. An investigation into the Winkler modeling of the cyclic response of rigid footings, Soil Dynamics and Earthquake Engineering, 28, 44–57. Anastasopoulos I., Gazetas G., Loli M., Apostolou M, Gerolymos N., 2010. Soil Failure can be used for Seismic Protection of Structures. Bulletin of Earthquake Engineering, 8, 309–326. Anastasopoulos I., Gelagoti F., Kourkoulis R., Gazetas G. 2011. Simplified Constitutive model for Simulation of Cyclic Response of Shallow Foundations: Validation against Laboratory Tests. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 137(12), 1154-1168. Anastasopoulos Ι. 2010. Beyond conventional capacity design : towards a new design philosophy. Soil–Foundation–Structure Interaction, Orense R.P., Chouw N., Pender M.J. (editors), CRC Press, Taylor & Francis Group : New York. Anastasopoulos I., Georgarakos T., Drosos V., Giannakos S., and Gazetas G. 2009b. Towards a reversal of seismic capacity design: Part B, Shaking-table testing of bridge pier-foundation system. Proceedings of the 3rd Greece-Japan Workshop on Seismic Design, Observation, and Retrofit of Foundations, National Technical University of Greece, Santorini, 407–419. Anastasopoulos I., Loli M., Georgarakos T., and Drosos V. 2013. Shaking Table Testing of Rocking−isolated Bridge Piers. Journal of Earthquake Engineering, 17(1), 1-32. Aoi S., Kunugi T., Suzuki W., Morikawa N., Nakamura H., Pulido N., Shiomi K., and Fujiwara H. 2011. Strong motion characteristics of the 2011 Tohoku-oki earthquake from K-NET and KiK-NET. SSA Annual Meeting, 2011. Apostolou, M., and Gazetas, G. 2005. Rocking of foundations under strong shaking: Mobilisation of bearing capacity and displacement demands. 1st Greece-Japan Workshop on Seismic Design, Observation, Retrofit of Foundations, 11–12 October, 2005, Athens, Greece. Apostolou M., Gazetas G., and Garini E. 2007. Seismic response of slender rigid structures with foundation uplifting, Soil Dynamics and Earthquake Engineering 27, 642–654. Bartlett P. E., 1976. Foundation Rocking on a Clay Soil. ME thesis, Report No. 154, School of Engineering, University of Auckland, New Zealand. Bienen B., Gaudin C., & Cassidy M.J. 2007. Centrifuge tests of shallow footing behavior on sand under combined vertical-torsional loading. Int. J. Physical Modeling in Geotechnics, 2, 1-21. Borja R.I., Wu W.H., Amies A.P., Smith H.A. 1994. Nonlinear lateral, rocking, and torsional vibration of rigid foundations. Journal of Geotechnical Engineering, ASCE, 120(3), 491–513. Borja R.I., Wu W.H., and Smith H.A. 1993. Nonlinear response of vertically oscillating rigid foundations. Journal of Geotechnical Engineering 119, 893–911. Bransby M.F., Randolph M.F. 1998. Combined loading of skirted foundations. Géotechnique, 48(5), 637-655. Butterfield R., Gottardi G. 1994. A complete three−dimensional failure envelope for shallow footings on sand. Géotechnique, 44(1), 181184. Chang B.J, Raychowdhury P., Hutchinson T., Thomas J., Gajan S. & Kutter B.L. 2006. Centrifuge testing of combined frame-wallfoundation structural systems. Proc. 8th US National Conference on Earthquake Engineering, April 18–22, San Francisco, CA, paper No. 998. Chatzigogos C.T., Pecker A., and Salençon J. 2009. Macroelement modeling of shallow foundations. Soil Dynamics and Earthquake Engineering 29(5), 765–781. Chen X.C., and Lai Y.M. 2003. Seismic response of bridge piers on elastic-plastic Winkler foundation allowed to uplift. Journal of Sound Vibration , 266, 957–965.

(b) On the contrary, seriously under-designed foundations limit the transmitted accelerations to levels proportional to their (small) ultimate moment capacity. This leads to much safer superstructures. In earthquake engineering terminology the plastic “hinging” moves from the columns to the foundationsoil system, preventing dangerous structural damage. (c) For tall-slender systems that respond seismically mainly in rocking, underdesigning the footings “invites” strong uplifting and mobilization of bearing capacity failure mechanisms. It turns out that the statically determined ultimate moment resistance is retained without degradation during cyclic loading, at least for the few numbers of cycles of most events  hence the geotechnical reliability in such a design. Moreover, the cyclic response of such foundations reveals that the amount of damping (due to soil inelasticity and uplifting−retouching impacts) is appreciable, if not large, while the system has a fair re-centering capability. These are some of the secrets of their excellent performance. (d) The key variable in controlling the magnitude of uplifting versus the extent of bearing−capacity yielding is the static factor of safety FS against vertical bearing−capacity failure. The designer may for example, choose to intervene in the subsoil to increase FS and hence enhance uplifting over soil inelasticity. Such intervention need only be of small vertical extent, thanks to the shallow dynamic “pressure bulb” of a rocking foundation. (e) In classical geotechnical engineering, avoiding bearing capacity failure at any cost is an unquestionably prudent goal. Seismic “loading” is different  it is not even loading, but an imposed displacement. Sliding mechanisms develop under the footing momentarily and hence alternatingly, and may only lead to (increased) settlement. It would be the task of the engineer to “accommodate” such settlements with proper design. The results and conclusions of this paper are in harmony with the numerous experimental and theoretical findings of Professor Bruce Kutter and his coworkers at U.C. Davis, and of Professors Alain Pecker and Roberto Paolucci and their coworkers in Paris and Milano.

11 ACKNOLEDGMENTS Τhe financial support for the work outlined in this paper has been provided through the research project “DARE”, funded by the European Research Council (ERC), “IDEAS” Programme in Support of Frontier Research. Contract/number ERC–2–9– AdG228254–DARE .

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Chopra A.K., and Yim C.S. 1984. Earthquake response of structures with partial uplift on Winkler foundation, Earthquake Engineering and Structural Dynamics, 12, 263–281. Crémer C., Pecker A., Davenne L. 2001. Cyclic macro-element for soil– structure interaction: material and geometrical nonlinearities. International Journal for Numerical and Analytical methods in Geomechanics, 25(12), pp. 1257–1284. Cremer C., Pecker A., and Davenne L. 2002. Modeling of nonlinear dynamic behaviour of a shallow strip foundation with macroelement. Journal of Earthquake Engineering 6, 175–211. Dobry R., and Gazetas G. 1986. Dynamic response of arbitrarily– shaped foundations, Journal of Geotechnical Engineering 113, 109–135. Drosos V., Georgarakos P., Loli M., Zarzouras O., Anastasopoulos I., Gazetas G. 2012. Soil–Foundation–Structure Interaction with Mobilization of Bearing Capacity : An Experimental Study of Sand. Journal of Geotechnical and Geoenvironmental Engineering (ASCE), 138(11), 1369-1386. Faccioli E., Paolucci R., and Vanini M., 1998. 3D Site Effects and SoilFoundation Interaction in Earthquake and Vibration Risk Evaluation. Final report of the European research project TRISEE, European Commission, Brussels, Belgium. Faccioli E., Paolucci R., and Vivero G., 2001. Investigation of seismic soil-footing interaction by large scale cyclic tests and analytical models. Proceedings of the 4th International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil Dynamics (S. Prakash, ed.), CD-ROM, S. Prakash Foundation publisher, San Diego, CA. Fardis M. N. (ed.) 2010. Advances in Performance-Based Earthquake Engineering. Springer , University of Patras, Greece, pp. 485. Federal Emergency Management Agency (FEMA), 2000. Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA-356, Washington, D.C. Figini R.. 2010. Nonlinear dynamic soil-structure interaction: Application to seismic analysis of structures on shallow foundations. Ph.D. thesis, Politecnico di Milano, Italy. Furumura T., Takemura S., Noguchi S., Takemoto T., Maeda T., Iwai K., Padhy S. 2011. Strong ground motions from the 2011 off-the Pacific-Coast-of-Tohoku, Japan (Mw=9.0) earthquake obtained from a dense nationwide seismic network. Landslides (available online, DOI: 10.1007/s10346-011-0279-3). Gajan S., Kutter BL. 2008. Capacity, settlement and energy dissipation of shallow footings subjected to rocking, Journal of Geotechnical and Geoenvironmetal Engineering, ASCE 134(8), 1129-1141. Gajan S., and Kutter B. L., 2009a. Contact interface model for shallow foundations subjected to combined loading. Journal of Geotechnical and Geoenvironmental Engineering 135, 407–419. Gajan S., and Kutter B. L., 2009b. Effects of moment-to-shear ratio on combined cyclic load-displacement behavior of shallow foundations from centrifuge experiments. Journal of Geotechnical and Geoenvironmental Engineering 135, 1044–1055. Garini E., Gazetas G., and Anastasopoulos I. 2011. Asymmetric ‘Newmark’ Sliding Caused by Motions Containing Severe ‘Directivity’ and ‘Fling’ Pulses. Géotechnique, 61(9), 753-756. Gazetas G. 1991. Formulas and charts for impedances of surface and embedded foundations. Journal of Geotechnical Engineering, ASCE, 117(9), 1363–81. Gazetas G., Anastasopoulos, I., and Apostolou, M., 2007. Shallow and deep foundations under fault rupture or strong seismic shaking. Chapter 9 in Earthquake Geotechnical Engineering, (K. Pitilakis, ed.), Springer Publishing, 185–215. Gazetas G., Apostolou M., Anastasopoulos I. 2003. Seismic Uplifting of Foundations on Soft Soil, with examples from Adapazari (Izmit 1999, Earthquake). BGA Int. Conf. on Found. Innov., Observations, Design & Practice, Univ. of Dundee, Scotland, September 25, 3750. Gazetas G., Mylonakis G. 1998. Seismic soil–structure interaction: new evidence and emerging issues, emerging issues paper. Geotechnique, Spec. Pub. ASCE, 75, 1119–74. Gazetas G. 1987. Simple physical methods for foundation impedances. Chapter 2 in Dynamics of Foundations and Buried Structures (P. K.

Benerjee and R. Butterfield, eds), Elsevier Applied Science, Barking Essex, UK, 44–90. Gazetas G. 1991. Formulas and charts for impedances of surface and embedded foundation. Journal of Geotechnical Engineering 117, 1363–1381. Gazetas G., and Apostolou M. 2004. Nonlinear soil-structure interaction: Foundation uplifting and soil yielding. 3rd U.S.-Japan Workshop on Soil-Structure Interaction, 29–30 March 2004, Menlo Park, CA. Gazetas G., and Kavvadas M. 2009. Soil–Structure Interaction. NTUA Publications, Athens, Greece. Gazetas G., Garini E., Anastasopoulos I. 2009. Effect of near–fault ground shaking on sliding systems. Journal of Geotechnical and Geoenvironmental Engineering 135, 1906–1921. Gelagoti F., Kourkoulis R., Anastasopoulos I., Gazetas G. 2012. Rocking Isolation of Low- Rise Frame Structures founded on Isolated Footings”, Earthquake Engineering and Structural Dynamics, 41, 1177-1197. Georgiadis M., and Butterfield R. 1988. Displacements of footings on sands under eccentric and inclined loading. Canadian Geotechnical Journal, 25, 199–212. Gerolymos N., Apostolou M., Gazetas G. 2005. Neural network analysis of overturning response under near-fault type excitation. Earthquake Engineering and Engineering Vibration, 4, 213–228. Gottardi G., Houlsby G.T., Butterfield R. 1995. The displacement of a model rigid surface footing on dense sand under general planar loading. Soils and Foundations, 35, 71–82. Gourvenec S. 2007. Shape effects on the capacity of rectangular footings under general loading. Géotechnique, 57(8), 637-646. Gourvenec S., Randolph M.F. 2003. Effect of strength non−homogeneity on the shape and failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique, 53(6), pp. 527-533. Harden C., Hutchinson T. 2006. Investigation into the Effects of Foundation Uplift on Simplified Seismic Design Procedures. Earthquake Spectra, 22 (3), pp. 663–692. Harden C.W., and Hutchinson T.C. 2009. Beam on nonlinear Winkler foundation modeling of shallow rocking–dominated footings. Earthquake Spectra, 25, 277–300. Houlsby G.T., Amorosi A., & Rojas E. 2005. Elastic moduli of soils dependent on pressure: a hyperelastic formulation. Géotechnique, 55(5), 383–392. Houlsby G.T., Cassidy M.J., Einav I. 2005. A generalized Winkler model for the behavior of shallow foundation. Geotechnique , 55, 449–460. Housner G. W. 1963. The behavior of inverted pendulum structures during earthquakes, Bulletin of the Seismological Society of America, 53(2), 403–417. Huckelbridge A.A., and Clough R. 1978. Seismic response of uplifting building frame. Journal of Structural Engineering, 104, 1211– 1229. Ishiyama Y. 1982. Motions of rigid bodies and criteria for overturning by earthquake excitations. Earthquake Engineering Structural Dynamics 10, 635–650. Kausel E., & Roesset J.M. 1975. Dynamic stiffness of circular foundations. J. Eng. Mech. Div., ASCE, 101, pp. 771–85. Kawashima K., Nagai T., and Sakellaraki D. 2007. Rocking seismic isolation of bridges supported by spread foundations. Proceedings of 2nd Japan-Greece Workshop on Seismic Design, Observation, and Retrofit of Foundations, Japanese Society of Civil Engineers, Tokyo, 254–265. Kirkpatrick P. 1927. Seismic measurements by the overthrow of columns. Bulletin of the Seismological Society of America 17, 95– 109. Knappett J.A., Haigh S.K., Madabhushi S.P.G. 2006. Mechanisms of failure for shallow foundations under earthquake loading. Soil Dynamics and Earthquake Engineering, 26, 91–102. Koh A.S., Spanos P., and Roesset J.M. 1986. Harmonic rocking of rigid block on flexible foundation. Journal of Engineering Mechanics 112, 1165–1180.

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Kourkoulis R., Gelagoti F., Anastasopoulos I. 2012. Rocking Isolation of Frames on Isolated Footings : Design Insights and Limitations. Journal of Earthquake Engineering, 16(3), 374-400. Kourkoulis R., Anastasopoulos I., Gelagoti F., Kokkali P. 2012. Dimensional Analysis of SDOF Systems Rocking on Inelastic Soil. Journal of Earthquake Engineering, 16(7), 995-1022. Kutter B.L., Martin G., Hutchinson T.C., Harden C., Gajan S., Phalen J.D. 2003. Status report on study of modeling of nonlinear cyclic load–deformation behavior of shallow foundations. University of California, Davis, PEER Workshop, 2003. Kutter B.L., Wilson D.L. 2006. Physical Modelling of Dynamic Behavior of Soil-foundation-superstructure Systems. International Journal of Physical Modelling in Geotechnics, 6(1), 1–12. Kutter B.L., Martin G., Hutchinson T.C., Harden C., Gajan S., and Phalen J. D. 2006. Workshop on modeling of nonlinear cyclic loaddeformation behavior of shallow foundations. PEER Report 2005/14, Pacific Earthquake Engineering Research Center, University of California, Berkeley, CA. Le Pape Y., & Sieffert J.P. 2001.Application of thermodynamics to the global modelling of shallow foundations on frictional material. International Journal for Numerical and Analytical Methods in Geomechanics, 25, 1377-1408. Luco J.E., and Westman R.A., 1971. Dynamic response of circular footings. Journal of the Engineering Mechanics Division, 97, 1381–1395. Makris N., and Roussos, Y. 2000. Rocking response of rigid blocks under near source ground motions. Géotechnique, 50, 243–262. Martin C.M., Houlsby G.T. 2001. Combined loading of spudcan foundations on clay : numerical modeling. Géotechnique, 51(8), 687-699. Martin G.R., and Lam I.P., 2000. Earthquake resistant design of foundations: Retrofit of existing foundations. Geoengineering 2000 Conference (GeoEng2000), 19–24 November 2000, Melbourne, Australia. Maugeri M., Musumeci G., Novità D., & Taylor C.A. 2000. Shaking table test of failure of a shallow foundation subjected to an eccentric load. Soil Dyn. and Earthq. Eng., 20 (5-8), 435-444. Meek J. 1975. Effect of foundation tipping on dynamic response, Journal of Structural Division, 101, 1297–1311. Mergos P.E., and Kawashima K. 2005. Rocking isolation of a typical bridge pier on spread foundation. Journal of Earthquake Engineering, 9(2), 395–414. Meyerhof G.G. 1963. Some recent research on the bearing capacity of foundations. Canadian Geotechnical Journal, 1(1), 6–26. Nakaki D.K., and Hart G.C. 1987. Uplifting response of structures subjected to earthquake motions. U.S.-Japan Coordinated Program for Masonry Building Research, Report No 2.1-3 (Ewing, Kariotis, Englekirk, and Hart, eds.). Negro P., Paolucci R., Pedrett S., and Faccioli E. 2000. Large-scale soilstructure interaction experiments on sand under cyclic loading. Paper No. 1191, 12th World Conference on Earthquake Engineering, 30 January–4 February 2000, Auckland, New Zealand. Nova R., & Montrasio L. 1991.Settlement of shallow foundations on sand. Géotechnique, 41(2), 243-256. Panagiotidou A.I., Gazetas G., and Gerolymos N. 2012. Pushover and Seismic Response of Foundations on Overconsolidated Clay: Analysis with P-δ Effects, Εarthquake Spectra, 28(4), 1589-1618. Panagiotidou A.I. 2010. 2D and 3D inelastic seismic response analysis of foundation with uplifting and P-δ effects. thesis, National Technical University, Athens, Greece. Paolucci R. 1997. Simplified evaluation of earthquake induced permanent displacements of shallow foundations. Journal of Earthquake Engineering 1, 563-579. Paolucci R. Shirato M., Yilmaz MT. 2008. Seismic behavior of shallow foundations : shaking table experiments vs. numerical modeling. Earthquake Engineering & Structural Dynamics, 37(4), 577-595. Paolucci R., and Pecker A. 1997. Seismic bearing capacity of shallow strip foundations on dry soils. Soils and Foundations 37, 95–105

Paulay T., and Priestley M.J.N. 1992. Seismic Design of Reinforced Concrete and Masonry Buildings. John Wiley & Sons, New York, NY. Pecker A. 2003. A seismic foundation design process, lessons learned from two major projects : the Vasco de Gama and the Rion Antirion bridges. ACI International Conference on Seismic Bridge Design and Retrofit, University of California at San Diego, La Jolla, USA. Pecker A. 1998. Capacity design principles for shallow foundations in seismic areas. Keynote lecture, in 11th European Conference Earthquake Engineering (P. Bisch, P. Labbe, and A. Pecker, eds.) A. A. Balkema, Rotterdam, The Netherlands, 303–315. Pender M. 2007. Seismic design and performance of surface foundations. 4th International Conference on Earthquake Geotechnical Engineering, Thessaloniki, Greece (CD-ROM). Priestley M.J.N. 1993. Myths and fallacies in earthquake Engineering―Conflicts between design and Reality. Bulletin, New Zealand Society for Earthquake Engineering , 26, 329–341. Priestley M.J.N. 2003. Myths and fallacies in earthquake engineering, revisited. Ninth Mallet-Milne Lecture, Rose School, IUSS Press, Instituto Universitario di Studi Superiori, Pavia, Italy. Raychowdhury P. & Hutchinson T. 2009. Performance evaluation of a nonlinear Winkler-based shallow foundation model using centrifuge test results. Earthquake Engineering and Structural Dynamics, 38(5), 679-698. Roesset J.M. 1980. Stiffness and damping coefficients of foundations, in Dynamic Response of Foundations: Analytical Aspects (M. W. O’Neil and R. Dobry, eds.). American Society of Civil Engineers, Reston, VA, 1–30. Salençon J., and Pecker A., 1995. Ultimate bearing capacity of shallow foundations under inclined and eccentric loads. Part II: Purely cohesive soil without tensile strength. European Journal of Mechanics, A:Solids, 14, 377–396. Shi B., Anooshehpoor A., Zeng Y., and Brune J. 1996. Rocking and overturning of precariously balanced rocks by earthquake. Bulletin of the Seismological Society of America 86, 1364–1371. Shirato M., Kouno T., Nakatani S., and Paolucci R. 2007. Large-scale model tests of shallow foundations subjected to earthquake loads, in Proceedings of the 2nd Japan-Greece Workshop on Seismic Design, Observation, and Retrofit of Foundations, Japanese Society of Civil Engineers, Tokyo, Japan, 275–299. Shirato M., Kuono T., Asai R., Fukui J., and Paolucci R. 2008. Large scale experiments on nonlinear behavior of shallow foundations subjected to strong earthquakes. Soils and Foundations, 48, 673– 692. Tassoulas J.L. 1984. An investigation of the effect of rigid sidewalls on the response of embedded circular foundations to obliquelyincident SV and P waves. Dynamic Soil–Structure Interaction, Rotterdam: A.A.Balkemal,. 55–63. Ticof J. 1977. Surface footings on sand under general planar loads, Ph.D. Thesis, University of Southampton, U.K. Ukritchon B., Whittle A.J., Sloan S.W. 1998. Undrained limit analysis for combined loading of strip footings on clay. Journal of Geotechnical and Geoenvironmetal Engineering, ASCE, 124(3), 265-276. Veletsos A.S., & Nair V.V. 1975. Seismic interaction of structures on hysteretic foundations. Journal of Structural Engineering, ASCE, 101(1), 109–29. Vesic A.S. 1973. Analysis of ultimate loads of shallow foundations. Journal of Soil Mechanics Foundation Div., ASCE, 99, 45–73. Vetetsos A.S., and Wei Y.T. 1971. Lateral and rocking vibration of footings. Journal of the Soil Mechanics and Foundation Division 97, 1227–1248. Wolf J.P. 1988. Soil–Structure Interaction Analysis in TimeDomain.Prentice–Hall, Englewood Cliffs, NJ. Zhang J., and Makris N. 2001. Rocking Response of Free-Standing Blocks Under Cycloidal Pulses. Journal of Engineering Mechanics, 127(5), 473–483.

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Figure 1. Conceptual illustration of (a) the response of a conventional and a “rocking-isolation” design of a bridge-pier foundation; and (b) the “capacity” design principle as conventionally applied to foundations, and its reversal in “rocking isolation”.

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Figure 2. Typical moment−rotation relations of three foundations and corresponding snapshots of their ultimate response with the contours of plastic deformation. The only difference between foundations : their static factor of safety.

Figure 3. Dimensionless Nu – Mu failure envelope for strip foundation

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Figure 4. Comparison of two slender systems (differing only in FS) subjected to monotonic and cyclic loading: (a) deformed mesh with plastic strain contours at ultimate state; (b) dimensionless monotonic moment–rotation response; (c) cyclic moment–rotation response; and (d) cyclic settlement– rotation response (the grey line corresponds to the monotonic backbone curves).

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Figure 5. (a) Two bridge piers on two alternative foundations subjected to a large intensity shaking, exceeding the design limits; (b) deformed mesh with superimposed plastic strain, showing the location of “plastic hinging” at ultimate state; (c) time histories of deck drift; (d) overturning moment−rotation (M−θ) response of the two foundations.

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Figure 6. (a) Two building frames on two alternative foundation subjected to a large intensity earthquake, exceeding the design limits; (b) deformed mesh with superimposed plastic strain, showing the location of “plastic hinging” at ultimate state; (c) bending moment–curvature response of the central columns; (d) overturning moment–rotation (M–θ) response of the two central foundations.

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Figure 7. (a) Old frame retrofitted with stiff Shear Wall on two different foundations  conventional B = 6 m and unconventional B = 3.5 m; (b) time histories on top floor drift ratio; (c) settlement–rotation curves of the Shear Wall footings.

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Figure 8. Sketches of damaged states of the three structures.

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Kerisel Kerisel Lecture lecture The role of Geotechnical Engineers in saving monuments and historic sites. Conférence Kerisel Le rôle des ingénieurs géotechniciens dans la sauvegarde des monuments et des sites historiques. Calabresi G. ISSMGE Technical Committee 301 core member

ABSTRACT: There are many interesting ways for geotechnical engineers to contribute to conservation issues. Firstly they can give a substantial contribution to the knowledge of the monuments and of their history. Then by assuming a broader, more comprehensive approach to the conservation issues, based on historical studies, possibly with the cooperation of scholars of different disciplines, they are often able to identify the nature, characteristics and evolution of the deterioration phenomena and to ascertain the necessity or opportunity of removing them. Eventually they can propose the less invasive solutions to save the monument and its material components that bear witness of its origin and history. The paper shows that in some cases this approach can be successfully applied to save historic buildings, while in others the origin of very slow soil movements, which increase the damage, can be very difficult to identify and furthermore costly investigations are required. However it is worthwhile to do any effort to achieve a convincing explanation of the distress causes and to propose interventions that are safe and respectful of the history of the monument. RÉSUMÉ : Les ingénieurs géotechniciens peuvent contribuer à la sauvegarde des anciens bâtiments et sites historiques de plusieurs façons. Premièrement, ils peuvent apporter une contribution importante à la connaissance des monuments et de leur histoire. Ensuite, avec une approche globale des questions de conservation, basée sur des études historiques, et éventuellement en coopération avec des chercheurs de différentes disciplines, ils peuvent souvent identifier la nature, les caractéristiques et l’évolution des phénomènes de dégradation et déterminer la nécessité ou la possibilité de les éliminer. Finalement, ils peuvent proposer les solutions les moins invasives pour sauver le monument et les matériaux qui le composent, témoins de son origine et de son histoire. Dans certains cas, cette approche peut être appliquée avec succès pour sauver les bâtiments historiques, tandis que dans d’autres, l'origine des mouvements du sol très lents peut être très difficile à identifier et les études nécessaires pour poursuivre les recherches deviennent très coûteuses. Néanmoins, il est justifié de faire tous les efforts possibles visant à parvenir à une explication convaincante des causes des désordres et de proposer des interventions qui soient sûres et respectueuses de l'histoire du monument. KEYWORDS: Geotechnical engineering, monuments, historic sites, conservation criteria, saving approach, underpinning, micropiles.

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ensuring such conservation. The complexity of the history and of the static and conservation conditions of historic buildings often generates problems in choosing the right intervention because of the presence of technical difficulties or because of differences in conservation criteria. All of this experience deserves being highlighted by promoting a critical discussion on the role of Geotechnical Engineers in saving monuments and historic buildings.

INTRODUCTION

Since the time when J. Kerisel (Kerisel 1975, 1987, 1997, 2004. Viggiani 1997, Isnard 1980) and Arrigo Croce (Croce 1980, 1985, Jappelli 1997) raised this issue, the theme of saving monuments and Historic Sites has gained interest and has seen an increasing involvement by geotechnical engineers. A contribution to this heightened interest has also come from the establishment and the activity of the ISSMGE Technical Committee (Tsatsanifos and Psarropoulos 2009) and from the impact of the debate that accompanied the search for solutions and the implementation of difficult interventions in the case of very famous monuments like the Tower of Pisa and the Cathedral of Mexico City. The theme is now a topical one in all Countries and often involves Geotechnical Engineers, but the close relationship between Geotechnics, history and evolution of engineering and architecture is particularly evident in Italy where towns, buildings and monuments built over a time period spanning thirty centuries, that are concrete evidence of how civilization has evolved in the Mediterranean, pose daily problems to restorers and conservation experts. In Italy, almost all buildings, monuments and historic sites have undergone successive changes throughout the centuries. Their history bears witness to the succession of events, interests, artistic trends, visions and to the evolution of construction techniques that have occurred over time. Their conservation demands contributions not only by the scholars of the Arts and Humanities, but also by technical experts who are capable of

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THE TECHNOLOGICAL PROGRESS OF GEOTECHNICAL ENGINEERING

It is self-evident that since all buildings interact with the ground on which they rise and are conditioned by its behaviour, their state of conservation is affected by any deformation of the soil and by any changes in its properties occurring naturally over time or caused by variations in environmental conditions. A monument, its foundation and the supporting ground should be considered as parts of a comprehensive complex system, that any saving proposal should take into account, but the soil is generally more sensitive than construction materials to stress variations and weathering; hence it is only natural that Geotechnics should be involved in discussions on saving and restoration problems since it is the discipline that more than any other investigates the nature and causes of soil displacements, and is therefore the best suited to finding ways of preserving ancient buildings and monuments.

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progressive instability caused by the erosion of the sand levels and by the excavations made during the Middle Ages, but also to free the stone face from the debris produced by the collapsed rock and discover the unsuspected presence of Hypogeums (Tomei and Filetici 2011). New intervention arose also from the progress achieved in the last decades in the knowledge of the behaviour of unsaturated soils and in the measurement of soil suction. Actually many old buildings with shallow footings suffer the effects of the shrinkage and swelling of unsaturated cohesive soils. The climate changes which occur in some world areas or the water level decrease produced by intense pumping lead often to new unattended settlements. However, as it has been recently proposed and implemented, control system of the saturation degree of the foundation soil can be carried out by means of subsurface porous water pipes, to be driven according to prearranged profiles (Carbonella et. al 2011).

The possibilities offered in this field by technological progress in Geotechnical Engineering in recent years have stimulated these activities all over the world, as is shown by the reports published in journals and in conference proceedings. Of course the potential of the new technologies opens up fascinating prospects in this sector; suffice it to think of the possibilities of introducing structural elements of any size into the soil or of mixing the soil with cement to turn it into a new coherent material that is very similar to concrete, or of injecting hardening materials that replace pore pressure fluids in predetermined points of the subsoil, using probes of all lengths that can travel in any direction, even along predetermined and controlled, curved lines. Actually, scientific progress and the great potential and flexibility of Geotechnical Engineering technology have allowed for the conservation and protection of important historic sites threatened by instability, landslides and weathering of the soils on which they rise. Suffice it to mention the measures taken to protect Orvieto, Italy (Fig. 1), that took more than a decade, with the anchoring of the high cliff faces made of soft pyroclastic rock (tuff) whose stability had been undermined by the slow softening of the Pliocene overconsolidated clays, present at their base (Manfredini et al. 1980, Martinetti 1981, Lembo Fazio et al. 1984, Tommasi et al. 1997, Tommasi and Ribacchi 1997, Pane and Martini 1997, Tommasi et al. 2005, Soccodato et al. 2013)

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b) Figure 2. a) The limestone cliff over the hermitage of Santa Caterina del Sasso. b) The anchoring consolidation works.

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A MORE RESPECTFUL APPROACH: PRESERVING THE KNOWLEDGE OF THE PAST

Quite often, for most engineers, the interaction between Geotechnics and the protection of ancient buildings is seen only from the standpoint of the design and execution of consolidation measures. First of all it has been noticed that measures taken to improve the static behaviour or seismic resistance of ancient buildings have not always had lasting effects, but on the contrary they have often produced even greater and irreversible damages. One example speaks for all: the Minaret of Mosul, Iraq, UNESCO Heritage monument (Fig. 3). The heavy, invasive, structural consolidation (by means of iron nails) and underpinning (micropiles) carried out in the 1981 (Lizzi, 1982, 1997) have not protected the monument from a further worsening of its static conditions, so much so that new

b) Figure 1. Orvieto: the tuff high cliff (a) consolidated by means of passive anchors, nails and drain pipes (b). Scheme of the strengthening works along the edge of the Rock (Cencetti et al., 2005).

Not as extensive but not less important are the anchorage works on another cliff face overlooking Lake Maggiore thanks to which the historic hermitage of Santa Caterina del Sasso (Fig. 2) has been saved (Balossi Restelli 2009). More recently, a set of fiberglass tie-rods and a masonry underpinning have stopped the collapse of large tuff blocks from the NW face of the Palatine Hill. This measure not only made it possible to stop the

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measures are most urgent, but it is extremely difficult and problematic to decide on how to go about such measures.

a) b) Figure 5. Some of the intervention measures proposed to save the Tower of Pisa at the 1973 call for projects (Burland et al. 2013): a) Fondedile proposal; b) Impredit-Gambogi-Rodio proposal.

Actually until the early 1990s, the concept that the conservation of a monument involves also saving its construction components, even those that are not visible had not yet gained ground; the idea that the Tower of Pisa, once it were to be transferred onto a new foundation built using the technologies of the 20th century, would become a fake, only a pure icon of the monument, was not understood (Calabresi and Cestelli Guidi 1990, Calabresi 2011). The new way of thinking made its way gradually and radically changed the cultural approach to the consolidation of ancient buildings, and in the case of the Tower of Pisa, it led to the solution that was finally and happily adopted for its stabilization (Burland et al. 2000).

Figure 3. The Minaret of Mosul, underpinned micropiles and structurally strengthened in 1981 (Lizzi 1982, 1997).

The role of Geotechnical Engineers in the conservation of historic towns and monuments could be much broader and multifaceted and even more attractive in cultural terms than what is generally believed. The general perception of geotechnical engineering only as a means for intervening in a historic structure from the static standpoint is restrictive and far from the present view of thinking about monument conservation. Indeed it is now common thinking that the replacement or substantial modification of a structure or of a foundation alters or even eliminates forever an historically essential feature of a monument, the idea being that even its non visible parts, like the foundations, must also be preserved as a material token of its history. A self evident example of the changing of mind that occurred in the course of a few decades is provided by the Leaning Tower of Pisa: for a long time, faced with the objective difficulty in interpreting the phenomena that were causing the progressive inclination of the Tower, technological solutions were offered that were intended to make the Tower independent of the behaviour of its foundation soil. In 1962, F. Terracina, a geotechnical engineer who was a passionate scholar of the Tower, published a proposal (Fig. 4) that simply envisaged the removal of soil from the uphill section (anticipating the solution adopted 40 years later) (Terracina 1962), but its suggestion remained unattended.

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THE NEED OF MULTIDISCIPLINARY STUDIES

If the protection of a historic and monumental building has the aim of maintaining and spreading the knowledge of past eras and civilizations, then the study of the interaction between buildings and the environment, and in particular their foundation soils, brings a substantial contribution to it; it may help understand the choices made by the designers at the time of construction, the changes that occurred over the years, the causes of damages, and the techniques and materials used and relate them to the natural and artificial materials available, to the machines and to the historic context. All this helps deepen our knowledge of remote times. In this setting the contribution offered by Geotechnics, alongside that offered by structural engineers, geologists, seismologists, architects, art historians and construction historians may play an extremely important role. The examples of activities carried out with this spirit are now a great many and have been quite successful with at times unexpected and surprising results. More than thirty years ago the archaeologist Gullini had already presented a fascinating picture of the results achieved through cooperation between geotechnical engineers, archaeologists and historians in studying the developments in construction techniques and design in antiquity (Gullini 1980). They studied the foundations of ancient monuments and archaeological settlements in Mesopotamia and in the Mediterranean area from the 4th millennium B.C. to the late Roman Empire. Today there are many conservation projects sponsored by UNESCO which have a multidisciplinary approach in which Geology and Geotechnics play an essential role: for instance mention can be made of the set of measures proposed for Greece presented by IAEG (Christaras 2003). An Italian example is the Valley of the Temples in Agrigento (Croce et al. 1980.): studies carried out on the slope stability of the area where the temples rise have contributed to a better understanding of the history of Magna Greece and of the technical culture of its inhabitants between the 6th and 5th centuries B.C. within the frame of our knowledge of ancient Greece architecture (Dinsmoor 1975).

Figure 4. Layout of the underexcavation proposed by Terracina (1962).

Geotechnical Engineering had made great progress (with the development of micropiles and consolidation techniques) and the call for projects launched to save the Tower in 1973, after the completion of the studies on its subsoil (Cestelli Guidi et al. 1971) attracted only projects that aimed at creating a deepseated underpinning (Fig. 5), across soils that were more or less deformable (Burland et al. 2013).

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to 1319, has a total height of 89 m; presently its axis has an inclination of 1°16' against the vertical, that is not increasing. In recent years there has been a widespread concern about the possible seismic vulnerability of the tower and an in-depth research has been carried out on its static and dynamic equilibrium conditions (Lancellotta, 2007, 2013). The main problem was whether the tower had a sufficient stability factor against a seismic action of assumed intensity.

a)

b)

Figure 7. The Cathedral and the Ghirlandina tower at Modena. A view of the leaning tower and the Cathedral apse.

c) Figure 6. The Temple of Juno at Akragas (Agrigento). a) The calcarenite cliff; b) An aerial view; c) An outline of its foundations Cotecchia et al. 2000).

The rational layout of the Greek town, Akragas, is only one of the many discoveries made (Fig. 6). Actually it is clear that the designers took into account the geomorphological characteristics of the area and they adopted solutions for the foundations that contemplated the properties of the soils and the seismic nature of the area (Cotecchia 1997, Cotecchia et al. 2000). Indeed, the foundations of the structural elements of all the temples, consisting of large calcarenite blocks were placed on the rigid and resistant calcarenite layer located at several metres depth, underneath the Pliocene outcrop of a medium hard clay: the foundation of the temple of Hera Lacinia is located at more than 7 metres below ground surface. Does this mean that the Greeks knew about the local amplification of seismic action induced by the clay layer? The ruins of Jupiter’s Temple, that had been built previously and that had collapsed before its completion, suggest that this may be the case. Being acquainted with all the details of a monument’s history is essential in studying how to conserve it and in finding the best measures to ensure its conservation without undermining its original characteristics. The recent study of the static condition of the leaning tower Ghirlandina in Modena (Figs. 7, 9) is a beautiful, outstanding exemplary demonstration of the importance of deep historic knowledge for explaining the nature and origin of the damages and of the effective contribution offered by a thorough geotechnical investigation. The Ghirlandina, that was designed by Lanfranco, a famous medieval architect, and built from 1099

Figure 8. The planimetric positon of the various historic buildings and of the ancient Roman road Aemilia (Lancellotta 2013).

The geotechnical characteristics of the site are very complex. Actually the foundation soil is a succession of geologically recent alluvial deposits, covered by a thick (more than 6 m) layer of ancient, man made heterogeneous landfills. The upper horizons down to about 22 m are formed by medium to high plasticity inorganic clays, with an abundance of thin laminae of sand and peat. The geological, geotechnical and geophysical investigations showed that various periods of emersion during the deposition of the thick alluvial deposit generated a series of layers overconsolidated by desiccation.

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In this connection another emblematic and famous case is the Cathedral of Mexico City (Ovando-Shelley et al. 1997, Tamez et al. 1997, Santoyo and Ovando-Shelley 2000, OvandoShelley and Santoyo 2001). Historic information made it possible to identify the origin of the differential settlements of the foundation soil, part of which had been consolidated by preColumbian works, and to design sub-excavation and soil consolidation measures to offset the differential settlements (Fig. 10). On the other hand, the studies on foundations have contributed to a thorough understanding of the historic events of the Cathedral and of the surrounding area.

A detailed history of the tower and the nearby Cathedral (Labate 2009), their original design and subsequent modifications, was obtained from the study of many archive documents and was checked against the comparison of the material and stylistic characteristics of the various masonry levels of both buildings. In addition, on the basis of archeological escavations made in 1913 (Sandonnini, 1983) and more recent investigations (Labate, 2009), it was possible to identify the position of the late medieval cathedral, the preLanfranco cathedral and the actual Lanfranco cathedral (Fig. 8).

Figure 9. Vertical sections of Ghirlandina tower: from the left, view towards West, view towards North, view towards South, view towards East (Lancellotta 2013).

Figure 10. Underexcavation at the Cathedral of Mexico City (Santoyo and Ovando-Shelley, 2000).

Since the foundation soil has “memory” of the previous loading history, this detailed reconstruction was the key to explain the differential settlements, suffered by the cathedral and in particular the tilt of its apse towards East and not only towards the Ghirlandina tower. Additional borings allowed to identify a detailed profile of the soil upper layer and to find the remains of the ancient Roman road Via Aemilia at a depth of about 7 m. By comparing the different elevations of its pavement below the tower and outside, it became possible to deduce the settlements of the tower and the compressibility of its foundation soil. In order to explore the stability equilibrium of the leaning tower (Cheney et al. 1991, Di Tommaso et al. 2012) the inverted pendulum model has been adopted. Its parameters were derived from the soil investigations and from an experimental identification analysis of the tower dynamic behaviour in the presence of ambient vibration. The model parameters were chosen according to the time histories of the tower vibration, collected by means of a set of accelerometers at different heights; then a thorough analysis of soil-structure interaction was carried out in order to get a reliable estimate of the rotational stiffness and of the dynamic response of the tower foundation. The results gave reason for the good performance of the tower during the past seismic events and showed that there is no need for underpinning interventions. Furthermore it appeared that if the tower had been underpinned on micropiles, following the dogmatic trend of 2030 years ago, the decrease of the fundamental period of the structure would have increased its seismic vulnerability.

5

CRITICAL CASES

There is a long list of monumental buildings that, owing to the slow or very slow displacements in the foundation planes, suffer progressive instability. In these cases a conflict sets in between the purely technological approach (aimed at reinstating the safety of the monument with structural interventions which, while ensuring that the external aspects are preserved, modify the original structural design), and a softer approach, on the other hand, that begins with a study of the phenomena underlying the instability and makes a long and perhaps uneventful search of the causes that need to be removed to stop the instability and if possible save the monument without substantial alterations so as to respect its historic integrity. It is worth recalling that the search for the causes is always a timeconsuming exercise that is often much more expensive than ordinary, obvious structural and geotechnical engineering interventions. A systematic study of the saving projects carried out in Italy until 1995, including buildings of different kinds (Table 1), has shown that pure underpinning by micropiles was the largely predominant type of measure (Fig. 11) which in many cases was probably unnecessary or unsuited.

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mere hypotheses: the Basilica of St. Angelo in Formis and the Pienza Cathedral. St. Angelo in Formis St. Angelo in Formis is a Benedictine basilica near Capua which rises on the slopes of a rock hill (Fig. 12); it was built in the 6th century A.D. on the ruins of a Roman temple whose origins date back to the 5th century B.C (Cammarota, 2013). The basilica, which has three naves, presents traces of the changes it underwent in time. In particular the bell tower and the portico probably collapsed and were rebuilt in the 13th century. The foundations of the apse, most of the external walls and the pillars of the naves are rather shallow and rest on a fractured dolomite mass, whereas the foundations of the facade, the portico and a small proportion of the side walls rest on debris deposits and backfill. The geology of the area is complex because the dolomite mass overlies more recent Oligocene and Myocene deposits and there are major fractures of tectonic origin (Fig. 13). There is knowledge of relevant repair and consolidation measures adopted in 1732 and in 1930 after seismic damages. Of the more recent earthquakes of 1962, 1970 and 1980, only the last one caused some slight damages. From the end of the 1960s some cracks of static origin appeared in the walls of the naves lying over the pillars and with their slow progression they have caused quite some alarm and have required underpinning props.

Table 1. Monuments types subjected to systematic study (from Cecconi et al. 1997).

Figure 12. The Benedictine Basilica of St. Angelo in Formis.

a) a)

b) Figure 11. An analysis of some Italian monuments (modified from Cecconi et al. 1997): a) damage types; b) preservation measures.

b) Figure 13. St Angelo in Formis. a) The main fissures; b) Geologic section of the foundation soil (Cammarota et al. 2013).

However sometimes the causes of the instability are not clear and the possibility of removing them remains at best uncertain. This is the case of two Italian monuments of great value for which, after years of investigations, the causes of their instability still have not been found and for which there are only

The geological and geotechnical investigations performed so far in different stages have not helped to identify the causes of the settlements of the foundation soil. A first hypothesis attributed the instability to the mining activities carried out using explosives in a nearby quarry, but even after the mining

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and the sandstone scarp that delimits it towards the Orcia Valley (Fig. 17): the apse, with its underlying crypt, had its foundations downhill from the scarp where the level of the ground is about 15 m lower (Fig. 15). The construction of the apse ran into considerable and unexpected difficulties. In his memoirs the Pope wrote that the foundation plane rested on rock masses crossed by large fissures and that large arches were built across them to support the foundations. Some fissures appeared in the walls of the apse before the completion of the construction works, but Rossellino attributed them to the setting of the mortar (Piccolomini 2008). The church was inaugurated on 29 August 1462. New cracks appeared soon after between the nave and the apse and in the underlying crypt. Since then, for five centuries, there has been an uninterrupted succession of instability phenomena and consolidation works under the foundations; drifts and deep drainage wells have been driven, reinforcement buttresses have been built to uphold the apse, repairs and restructuring measures have been adopted for the side walls, the crypt under the apse, the vaults and the roof (Di Pasquale 1992). All these measures were made necessary by the constant lowering of the apse foundation downhill from the rock scarp: there is proof that between 1520 and 1530 the floor of the apse was al-ready lower than that of the nave by about 27 centimetres. A sudden settlement of about 0.3 m of the soil downhill from the scarp occurred on the night of 26 November 1545 and caused the partial collapse of the apse and of the bell tower. The event, described in the memoirs of a citizen of Pienza is defined Terrae motus (literally a movement of the earth), but there are doubts about it being an earthquake or a sudden slope instability phenomenon, perhaps triggered by a seismic quake. At present the overall difference in level of the apse with respect to the nave is about one metre, as it can be seen from the relative displacement of the cornice in Figure 16.

activities stopped in 1981 the cracks and fissures continued to widen. The origin of the distress remains unclear so that further geotechnical investigations and more extended studies are necessary. The safety of the fissured masonry structures – arches and vaults – is currently ensured by provisional and removable props, but while it is increasingly difficult to obtain public economic support to carry out research into the causes of the on-going phenomena, the proposals of consolidating the masonry walls of the basilica by means of important structural measures are bound to increase. The Pienza Cathedral The Pienza Cathedral (Fig. 14) is perhaps less famous than the Tower of Pisa, but it is just as problematic and intriguing.

Figure 14. Cathedral of Pienza and Piccolomini Palace from the square.

Perhaps there is no other monument that, in its lifetime, has been subjected to so many consolidation and strengthening measures as the Pienza Cathedral, because of the very slow, but continuous settlements of the foundation soil underneath its apse (Forlani Conti 1986).

Figure 16. The cornice displacement shows the apse settlement.

Figure 15. The Piccolomini Palace and the Pienza Cathedral apse seen from the rock scarp downhill.

In 1459 Enea Silvio Piccolomini, newly elected pope with the name of Pius II, decided to raise the status of his birth town with the construction of a Cathedral and some noble palaces. Works for the construction of the new cathedral started in 1459 and were completed in only three years. In order to make sure that the cathedral would be of appropriate proportions without restricting the size of the main square, situated symbolically between the Cathedral and the City Hall, the architect, Bernardo Rossellino extended its layout beyond the walls of the village

Figure 17. Planimetric position of the Cathedral.

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Main events in the Cathedral history (Fig. 18) 1458, Enea Silvio Piccolomini is elected pope (Pius II) and begins to rebuild his home town. 1459, The construction of the Cathedral is initiated by Bernardo Rossellino. Some problems arise in the apse foundation. 1462, At the end of the works some fissures appear close to the first choir chapel. Their openings increase towards the vaults. (a) 1462, Aug. 29, Solemn opening of the Cathedral. 1490, The apse begins to settle. 1500, The settlement of the apse reaches about 0.3 m. 1503, A tunnel is built to drain water from under the crypt. 1508, Two buttress piers are built against the apse to sustain it.(b) 1514, A wall is built around the apse. The settlement reaches 0.45 m 1545, Nov. 25, An earthquake (?) causes a large settlement of the apse and the collapse of the belfry; a crack appears along the natural scarp, SW of the town. (c) 1570, Repair works of the earthquake damages are completed. The collapsed vaults of the transept are substituted by false vaults 1596, The crypt arches and the external walls are streightened. 1604, A. Sandrini, architect, states that the damages are due to the slope movement, so that underpinning the apse is useless. (d) 1650 - 1760, Repeated repair works. The apse is more than 25 cm out of the plumb line. 1750 - 1770, A proposal of demolishing the apse and shortening the church is considered, but happily not carried out. 1888 - 1895, Collapsing vaults are replaced by false works. The pillars are connected by steel tendons. The apse has settled 0.85 m and has increased its detachment from the nave. (e) 1911 - 1925, Underpinning of the apse with masonry pillars, which bypass the sandstone layer to reach the marly clay. (f) 1926 - 1929, Various repair works are carried on, the apse walls are strengthened. The transept is underpinned. 1930 - 1933, The apse and crypt vaults are rebuilt. (g) 1958 - 1962, Underpinning of nave and aisle pillars with root piles. A hydraulic diaphragm is built around the front. (h)

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engineer who identified two traces of a vicarious fault, practically aligned with the cracks in the walls, having a total throw of about 15 m. Quite surprisingly the new structural consolidation measures adopted a few years later did not take this fact into any account. Later, between 1979 and 1984 a more thorough survey was made of the structures and of the relevant instability and geological and geotechnical investigations were carried out to define the bedding and mechanical characteristics of the foundation soils. The following figures (Figs. 19, 20) show two stratigraphic sections and their positions in the plan view. Under the square and the nave of the Cathedral a limestone layer 3 to 4 m thick over-laps the weakly cemented sand and fissured sandstone layer, having a thickness of 12-15 metres, which can be seen in the scarp on the sides of the apse (Lazzarotto and Micheluccini 1986, Calabresi et al. 1988, Calabresi et al. 1998).

g)

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h) Figure 18. Main events in the Pienza Cathedral history: a) 1462; b) 1503-1508; c) 1514-1545; d) 1570-179e) 1888-1895; f) 1908-1920; g) 1922 -1933; h) 1955 - 1979 (Di Pasquale 1992).

The phenomenon has always been attributed to the poor quality of the foundation soil, to its many fissures and to the effects of underground water. The sole exception is a report on instability dated 1604, in which an architect, A. Sandrini, having noticed that the earth surface fissures caused by the 1545 displacements were aligned with the scarp and extended throughout the whole southern side of the village, stated that the apse settlement was due to the movement of the slope; this interpretation of the phenomenon has been systematically ignored. In about 1750, as the instability in the area of the apse continued, suggestions were made to demolish that part and restrict the Cathedral to the part rising on the uphill part of the scarp. Luckily the proposal was not followed up and further measures were taken to consolidate the walls and foundations of the apse. In 1911, as concerns grew for the stability of the Cathedral, a lively debate occurred between those who suggested underpinning the apse and those who, following the example of what had just be done for the Spina Church in Pisa, suggested dismantling the apse to build a new foundation. In any case everyone was persuaded that the settlement of the apse was due to the poor quality of the foundation soil. Luckily the first position prevailed and between 1911 and 1929, by means of sample excavations through the sandstones to the underlying marly clays, stone and brick pillars were built under the apse down to more than 20 m from ground level downhill from the scarp. It was deemed that the apse, provided with a rigid monolithic foundation resting on the layer of marly clays, was finally consolidated. However, cracks causing detachment of the apse from the nave occurred again quite soon and in 1956 a new study committee was appointed including a geologist

b) Figure 19. a) The Cathedral plan and the traces of the fault traces; b) Geologic section AA, parallel to the church axis;

Figure 20. The axial section BB shows the position of the foundation block relative to the fault planes.

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Then it was finally stated that the apse settlement is not due to the deformation of the foundation soil but to the constant lowering of the area downhill from the set of faults.

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Figure 21. The main fault surface about 400 m N-W of the Cathedral

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It lies on a Pliocene formation of strongly overconsolidated marly clays whose thickness is about 100 m at the centre of Pienza and increases to over 800 m southwards (Brogi et al. 2005). The substrate of the Pliocene sediments consists of Mesozoic, carbonate-siliceous formations of the “Tuscan Series”. The discontinuities that border the scarp, already identified in 1956, are a set of locally vicarious faults having a WNW-ESE direction and southward dip. They are crossed by minor, approximately perpendicular discontinuities. The geotechnical investigations showed that both the sandstones and the underlying marly clays have high strength and negligible compressibility. In 1983 a periodical levelling was started by installing many benchmarks, uphill and downhill from the scarp and from the set of faults (Fig. 22). The measurements, repeated every year until 1992 (Guidi 1986) then at various intervals between 1994 and 1999 and resumed recently, show that the whole area covered by the bench-marks downhill from the scarp has a constant non uniform settlement of between 1 and 2 mm per year (Figs. 23, 24). Minor effects of this phenomenon are visible in other buildings in the same area (Costantini and Lazzarotto 2010). The lack of uniformity of the settlement rate shows that the Pliocene marly clay is split by the sets of discontinuities; the main vicarious fault is the main, but not the only source, of the soil displacement downhill from the scarp.

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b) Figure 23. Settlements of two significant points close to the Cathedral apse: a) plan view; b) settlements vs. time

0.0 ‐ 0.4 mm/a 0.4 ‐ 0.8 mm/a 0.8 ‐ 1.1 mm/a 1.1 ‐ 1.5 mm/a > 1.5  mm/a

Figure 24. Settlement rate contours in the area south of the Cathedral.

Figure 22. Ground settlement contours from June 1983 to January 1992.

Horizontal displacements are null or non measurable. The steady and extremely small rate of the movement, detectable only by a high precision levelling over a long term campaign explains why the phenomenon was never detected in the past.

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At the present time the masonry block constituting the underpinning of the apse built at the beginning of last century, whose internal edge lies uphill from a fault plane, while the external part is downhill, has a rigid downhill rotation which involves the overlying apse. Since the existence of an active fault should be ruled out, the only hypothesis that would account for the continuous settlement is a deep seated gravitational slope deformation within the marly clay formation, influenced by the shape of its bed and by the discontinuity surfaces (Genevois and Tecca, 1984, Calabresi 1992, Calabresi et al. 1995, Calabresi et al. 1988, Sciotti and Calabresi 2004). A recent seismic investigation along a longitudinal section measuring more than 1000 metres has highlighted a significant anomaly in the P-wave velocity contours under the Cathedral apse and a depression in the bed of the Pliocene deposits in the zone where the surface movements are largest, thus confirming that the faults detected at the surface involve also the underlying

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Mesozoic formation (Fig. 25). The lines of larger Vp gradient obtained from the seismic reflection investigation (Fig. 26) show the main units of the stratigraphic section and the probable traces of the discontinuity surfaces (Brogi et al. 2003). A likely hypothesis is that the sinking of a dihedral mass between two convergent fault planes is made possible by a horizontal displacement rate of the downhill zone too small to be topographically detectable. While it is evident that the faults, along whose surface the clay shear strength has a residual value, and the sets of minor discontinuities have a critical influence on the equilibrium conditions of the slope, their geometric characteristics and the cleft water pressures (Calabresi and Manfredini 1973, Sciotti and Calabresi 2004) have not yet been sufficiently defined to get a convincing explanation of the phenomenon. The project of a deeper geostructural and geotechnical research has been recently submitted to the study committee recently charged of carrying out an updated analysis of the Cathedral conditions, but its implementation has been delayed by economic problems. However the fundamental question still remains: assuming that the above explanation be correct, could a geotechnical measure, such as a decrease of the piezometric head, be designed to slow down the movement?

opinions. Some structural solutions were presented and discussed at a special conference (Mascardi 1992, Migliacci 1992), where however the concept of protecting the monument and its history also from a material point of view, without modifying its original design with inappropriate changes, largely prevailed. Luckily the rates of subsidence and rotation of the apse are very small and leave time for geotechnical engineers to look for a possible soft solution. There is a hope that they may win the challenge as it has happened for the Tower of Pisa. CONCLUSIONS Geotechnics may offer a significant contribution to the knowledge of ancient designs and construction techniques and to the interpretation of the causes of instability. The effects of deformations in foundation soils that occurred in ancient times, or that are difficult to trace back to any specific cause, can often be observed in ancient buildings. An ancient building or historic site is interesting in and of itself for geotechnical engineers, since it constitutes a monitoring instrument of the long term behaviour of the soil that influences them. The progress of geotechnical engineering and of the specialized technologies offer the means to perform complex and efficient interventions to save monuments, historic buildings and old towns. However too often in the last decades the new opportunities offered by this progress and the cooperation of geotechnical engineers has been utilized inappropriately by applying new deep foundations and structural modifications, that overcome the ancient building distress in a simplistic way, that ignores the history of the object of the intervention, from its initial building to our time, and the witness value of the technical solutions adopted by our predecessors and of their expertise. The great challenge is how to save monuments and historic buildings together with the physical token of their conception, their original construction techniques and their historic modifications, that are tangible witnesses of the history of mankind. The problems posed by slow, continuous settlements induced by deep seated deformations, which require long, in-depth and expensive investigations, are among the most difficult to be understood and explained. However the geotechnical engineers should feel themselves engaged in exploiting their knowledge of soil mechanics and applied geology to look for a way, if it exists, to save monuments and historic sites by removing the cause of distress and avoiding heavy structural interventions that distort their substantial characters. The cooperation of architects, historians, archaeologists, structural and geotechnical engineers is the necessary precondition for a respectful attitude towards conservation problems. In this context the geotechnical engineers have also the opportunity of actively contributing to the knowledge of the history of architecture and engineering, by following the unforgettable example and the footsteps of our great colleague Jean Kerisel.

a)

b) Figure 25. Seismic refraction tomography. a) The section trace; b) P-wave velocity contours.

Figure 26. Wave P velocity gradients from the seismic reflection measurements. 1. Remoulded superficial soil; 2. Limestone and Sandstone (Pliocene); 3. Over-consolidated marly clays (Pliocene); 4, 5 carbonate-siliceous formations “Tuscan Series” (Mesozoic); 6 Anhydrite (Trias).

ACKNOWLEDGEMENTS The Author is very grateful to his colleagues of the Pienza Scientific Committee Antonio Lazzarotto and Silvia Briccoli Bati for their continuous, friendly collaboration and for their help in gathering the relative documentation. The very precious help of Dr. Manuela Cecconi in preparing the paper is gratefully acknowledged.

Consolidation measures of the Cathedral of an entirely different approach, aimed at supporting the apse area with new reinforced concrete structures hinged to the ground uphill from the fault, are repeatedly being submitted by groups with different

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013 Christaras B. 2003. Cultural Heritage and engineering geology factors of damage. Engineering Geological Factors of Damage at Greek Monuments and Sites included in the World Heritage List of UNESCO. 2, 37-55 Technica Chronica, Athens Costantini A. and Lazzarotto A. 2010 Pienza città rinascimentale: i dissesti del Duomo e del Centro Storico. Etrurianatura 7, 41-56. Cotecchia V. 1997. Geotechnical degradation of the archaelogical site of Agrigento. Proc. Int. Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 101-108 C. Viggiani ed., A.A.Balkema, Rotterdam Cotecchia V., L. Monterisi and S. Rana, 2000. Condizioni di stabilità e interventi di consolidamento del tratto di collina di Agrigento in corrispondenza del Tempio di Giunone Lacinia. Proc. Int. Symposium GeoBen. National Research Concil, Torino Croce A. 1980. Geotecnica e beni culturali. Atti XIV Convegno Nazionale di Geotecnica. Associazione Geotecnica Italiana. Firenze Croce. A et al. 1980. La città di Agrigento e la Valle dei Templi. Atti XIV Convegno Nazionale di Geotecnica. Associazione Geotecnica Italiana. Firenze Croce A. 1985. Old monuments and cities - Research and preservation. Geotechical Engineering in Italy. Associazione Geotecnica Italiana, Roma Dinsmoor W.B. 1975 The architecture of ancient Greece. W.W.Norton & Co, New York Di Pasquale S. 1992. Analisi della stabilità del monumento. . Il Duomo di Pienza, Cinque Secoli di Restauri. Atti Conv. Pontignano 37-44 Soprintendenza BB.AA.AA., Siena Di Tommaso A., Lancellotta R., Focacci F., Romano F. 2012. Seismic capacity of the Ghirlandina Tower in Modena. Int. Conf. on Structural Analysis of Historical Constructions, Jerzy Jasienko ed., Wroclaw, Poland, 1474-1484 Forlani Conti M. 1986. La Cattedrale di Pienza e i suoi dissesti attraverso i documenti di archivio. Il Duomo di Pienza, 1459-1984, Studi e Restauri. 17-57 Cantini, Firenze Forlani Conti M. 1992. Il Duomo di Pienza, Cinque secoli di restauri. Il Duomo di Pienza, Cinque Secoli di Restauri. Atti Conv. Pontignano 21-35 Soprintendenza BB.AA.AA., Siena Genevois R. and Tecca P.R. 1984. Alcune considerazioni sulle «Deformazioni gravitative profonde” in argille sovraconsolidate. Boll. Soc. Geol. It., 103.717- 729. Guidi F. 1986 Studi topografici con livellazione geometrica di alta precisione. Il Duomo di Pienza, 1459-1984, Studi e Restauri. 106109 Cantini, Firenze Gullini G. 1980. Le fondazioni e il supporto fisico del costruito nelle culture architettonoche dell'antichità. Atti XIV Convegno Nazionale di Geotecnica. Associazione Geotecnica Italiana, Firenze Isnard, S. 1990. Le Comité Technique 19 de la Société Internationale de Mévanique des Sols: Génie géotechnique et préservation du patrimoine culturel. The Engineering Geology of Ancient Works, Monuments Historic Sites. p 1965-70 Balkema, Rotterdam Izzo S., Lazzarotto A. and Menicori P. 1992 Elementi geologici dell’area di Pienza. Il Duomo di Pienza, Cinque Secoli di Restauri. Atti Conv. Pontignano 21-35 Soprintendenza BB.AA.AA., Siena Jappelli, R. 1997 An integrated approach to the safeguard of monuments: the contribution of Arrigo Croce. Proc. Int. Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 11-27 C. Viggiani ed., A.A.Balkema, Rotterdam. Jappelli, R. 1997 Rcommendations and prejudices in the realm of foundation engineering in Italy. A histoical review. Proc. Int. Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites.. 191-214 C. Viggiani ed., A.A.Balkema, Rotterdam. Kerisel, J. 1975. Old Structures in relation to soil conditions. Géotechnique, 25, 433-483 Kerisel, J. 1987 Down to earth: foundations past and present: the invisible art of the builder. Balkema, Rotterdam Kerisel, J. 1997 Geotechnical problems in the Egypt of Pharaos. Proc. Int. Symposium Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 33-40 C. Viggiani ed. A.A.. Balkema, Rotterdam Kerisel J. 2004 Pierres et hommes des Pharaons à nos jours. Presses de l'E.N.P.C., Paris Labate D. 2009. Il contributo dell’archeologia alla lettura di un monumento. La Torre Ghirlandina: un progetto per la conservazione. Vol. 1, 66-77, Luca Sossella Editore, Roma.

REFERENCES Associazione Geotecnica Italiana. 1991. The contribution of geotechnical engineering to the preservation of Italian historic sites. X European Conference I.S.M.F.E., Firenze Balossi Restelli A. 2009. Eremo di S. Caterina del Sasso. Interventi di consolidamento. Tecniche di miglioramento dei terreno Programma di formazione permanente. Politecnico di Milano Berti A., 1986 Indagini geomorfologiche e idrogeologiche. Il Duomo di Pienza, 1459- 1984, Studi e Restauri. 99-102 Cantini, Firenze Brogi A., Lazarotto A., Liotta D. and Ranalli G. 2003. Extensional shear zones as imaged by reflection seismic lines: the Larderello geothermal field. Tectonophysics, 363, 127 - 139 Brogi A., Lazzarotto A., Liotta D. and CROP18 Working Group. 2005. Structural features of southern Tuscany and geological interpretation of the CROP 18 Seismic Reflection Survey. Bollettino Società Geologica Italiana, 3, 213 - 236 Burland, J.B., Jamiolkowski, M.B., Squeglia N., Viggiani, C. (2013). The leaning Tower of Pisa. Second International Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites. Balkema Rotterdam (in print) Burland, J.B., Jamiolkowski, M.B., Viggiani, C. (2000). Underexcavating the Tower of Pisa: Back to the future. GEOTECH-YEAR 2000, Developments in Geotechnical Engineering, Bangkok,Thailand, Balasubramaniam,A.S. et al. Eds, pp. 273-282 Calabresi G. 1986 I terreni e le strutture di fondazione. Il Duomo di Pienza, 1459-1984, Studi e Restauri. 144-153 Cantini, Firenze Calabresi G. 1992. Le fondazioni del Duomo di Pienza. Il Duomo di Pienza, Cinque Secoli di Restauri. Atti Conv. Pontignano 16-21 Soprintendenza BB.AA.AA., Siena Calabresi G. and Manfredini G. 1973. Shear strength characteristics of the jointed clay of S. Barbara. Géotechnique, 23 (2), 233 –244 Calabresi G., A. Lazzarotto and M. Micheluccini. 1988. The Cathedral of Pienza and its foundation soils. The Engineering Geology of Ancient Works, Monuments Historic Sites. Balkema, Rotterdam Calabresi G. and Cestelli Guidi C. 1990. Le attuali ccondizioni di stabilità della Rorre di Pisa. Materiali e Strutture: Problemi di Conservazione I, n.1 L’Erma di Bretschneider, Roma Calabresi G., Izzo S., Lazzarotto A., Menicori P. & Pieruccini U. 1995. Movimenti gravitativi nell’area di Pienza. Boll. Soc. Geol. It. 50, 67-82 Calabresi G. and (calabresi and D'Agostino 1997, S. 1997. Monuments and historic sites: intervention techniques. Proc. Int. Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites. C. Viggiani ed., A.A.Balkema, Rotterdam. Calabresi G. et al. 1998. The Cathedral of Pienza and its foundation soils. Proc. Int. Symposium Engineering Geology of Ancient Works, Monuments and Historic Sites. Balkema, Rotterdam Calabresi G. 2011. The soft approach to saving Monuments and Historic Sites. Proc. XV European Conference I.S.M.G.E. Athens Calabresi G. 2011. Recupero e conservazione del costruito d'interesse storico e monumentale: aspetti geotecnici. ARCo - Progetti d'eccellenza per il restauro italiano. Gangemi Editori, Roma Cammarota A.,Russo G., Viggiani C., Candela M. 2013 The Benedictine Basilica of S. Angelo in Formis (Southern Italy): a therapy without diagnosis? Proc. 2nd Int. Symposium on Geotechnical Engineering for the preservation of monuments and historic sites, A.A.Balkema, Rotterdam in print Carbonella M, Cenni G., Franceschini M. 2011. Stabilizzazione di terreni argillosi soggetti a fenomeni di ritiro e rigonfiamenteo: un intervento eseguito su un fabbricato dissestato sito a Bologna. XIV CNG. 2, 367-374 Associazione Geotecnica Italiana, Roma Cecconi M., Croce P. and D'Amelio M.G. 1997. Comparative analysis of some Italian monuments. Proc. Int. Symposium on Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 69-80 C. Viggiani ed., A.A.Balkema, Rotterdam. Cencetti C., Conversini P. and Tacconi P. 2005. The Rock of Orvieto (Umbria, Central Italy). Giornale di Geologia Applicata 1, 103112. Cestelli Guidi, C., Croce, A., Skempton, A.W., Schultze, E., Calabresi, G., Viggiani, C. 1971. Caratteristiche geotecniche del sottosuolo della Torre. Ricerche e studi sulla Torre pendente di Pisa ed i fenomeni connessi alle condizioni d’ambiente, IGM, Firenze, I, pp. 179-200. Cheney, J.A., Abghari, A., Kutter, B.L. 1991. Leaning instability of tall structures. Journal of Geotechnical Engineering, ASCE, 117(2): 297-318.

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013 Lancellotta, R., Pepe, M., 1998. On the stability of equilibrium of the Leaning Tower of Pisa, Atti Sc. Fis. Accademia delle Scienze, Torino, 132, pp. 1-11. Lancellotta R. 2009. Aspetti geotecnici nella salvaguardia della torre Ghirlandina. La Torre Ghirlandina. Un progetto per la conservazione. p. 178-193 Luca Sassella Editore, Roma, Lancellotta R. 2013. La Torre Ghirlandina: una storia di interazione struttura-terreno. XI Croce Lecture, Rivista Italiana di Geotecnica, in print Lazzarotto A. & Micheluccini M., 1986. Indagini geologiche. Il Duomo di Pienza, 1459-1984, Studi e Restauri. 94-98 Cantini, Firenze Lembo Fazio A., Manfredini G., Ribacchi R., Sciotti M. 1984 Slope Failures and Cliff Instability in the Orvieto Hill Proc. 4rd Int. Symp. on Landslides 2, 115-120, Toronto Lizzi F. 1982. The static restoration of monuments. Sagep Editice, Genova Lizzi F. 1997. The Pali Radice (Mcropiles) for the preservation of monuments and historic sites. Proc. Int. Symposium Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 463-469 C. Viggiani ed., A.A.Balkema, Rotterdam. Manfredini G., Martinetti S., Ribacchi R., Sciotti M., 1980. Problemi di stabilità della Rupe di Orvieto. XIV Conv. Naz. di Geotecnica, 1, 231-246, Associazione Geotecnica Italiana, Firenze Martinetti S. 1981. Saving old towns on hill top. Proc. X ICSMFE, Stockholm 4, 841-846 Mascardi C. 1992 Ipotesi di consolidamento strutturale e di restauro. Il Duomo di Pienza, Cinque Secoli di Restauri. Atti Conv. Pontignano 69-83 Soprintendenza BB.AA.AA., Siena Migliacci A. 1992 Ipotesi di consolidamento strutturale e di restauro. Il Duomo di Pienza, Cinque Secoli di Restauri. Atti Conv. Pontignano 69-83 Soprintendenza BB.AA.AA., Siena Ovando-Shelley E. and Santoyo E. 2001 Underexcavation of buildings in Mexico City: the case of the Metropolitan Cathedral and The Sagrario Church Proc. ASCE Journal od Architectural Engineering Ovando-Shelley E., Tamez E and Santoyo E. 1997 Geotechnical aspects for underexcavating Mexico's City Metropolitan Cathedral: main achievements after three years. Proc. Int. Symposium Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 479-488 C. Viggiani ed., A.A.Balkema, Rotterdam. Pane V. and Martini E.1997 The preservation of historical towns in Umbria: The Orvieto Case and its observatory. Proc. Int. Symposium Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 489-498 C. Viggiani ed., A.A.Balkema, Rotterdam. Piccolomini E.S. 2008. Commentarii (Latin with Italian translation, L. Totaro ed). 1576, 1744-1771 Adelphi Edizioni, Milano

Sandonnini T. 1983. Cronaca dei restauri del Duomo di Modena (18971925), a cura di O. Baracchi, Aedes Muratoriana, Modena 285 pp. Santoyo E.V. and Ovando-Shelley E. 2000. Mexico's City Cayhedral and Sagrario - Geometrical correction and soil hardening. Mexico City's Cathedral and Sagrario, TGC Ingenieria, Mexico Sciotti. A. and Calabresi G. 2004 Deep-seated movements in stiff jointed clays: the role of structural discontinuities”. Advances in geotechnical engineering: the Skempton Conference. Thomas Telford, London. Soccodato F.M, E. Martini, L. Tortoioli and A.M. Mazzi. 2013. The preservation of historical, archaeological and artistic heritage of Orvieto: an interdisciplinary project. Proc. 2nd Int. Symposium on Geotechnical Engineering for the preservation of monuments and historic sites. Associazione Geotecnica Italiana, in print Tomei M.A. and Filetici M.G. (eds) 2011 Domus Tiberiana - Scavi e Restauri. Electa, Roma Tamez, E., Ovando-Shelley E., Santoyo E., 1997 . Underexcavation of the Metropolitan Cathedral in Mexico City Proc. XIVth ICSMFE, 4, 2105-2126Hamburg Terracina F. 1962. Foundations of the tower of Pisa. Géotechnique 12 (4) 336-339 Tommasi P., R. Ribacchi and M. Sciotti 1997. Geotechnical aspects in the preservation of the historical town of Orvieto. Proc. Int. Symposium Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 849-858 C. Viggiani ed., A.A.Balkema, Rotterdam. Tommasi P., Ribacchi R., 1998. Mechanical behaviour of the Orvieto tuff. 2nd Int. Symp. Hard-Soils and Soft-Rocks 2, 901-909, Napoli Tommasi P., Boldini D., Ribacchi R., 2005. Twenty-year monitoring of the Orvieto overconsolidated clayey slope (Italy). XVI International Conference on Soil Mechanics and Geotechnical Engineering, 2595-2598 Osaka. Tsatsanifos C. and Psarropoulos P. 2009. TC 19 Preservation of Historic Sites – Administrative Report, Proc. XVII ICSMGE, 3763-64 Osaka Viggiani C. 1997. Laudatio Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 29-32 A.A.Balkema, C. Viggiani ed., Rotterdam. Viggiani C. 1997. Opening Address Geotechnical Engineering for the Preservation of Monuments and Historic Sites. 3-9 C. Viggiani ed., A.A.Balkema, Rotterdam Viggiani C. 2013 Cultural Heritage and Geotechnical Engineering: an introduction. Proc. 2nd Int. Symposium on Geotechnical Engineering for the preservation of monuments and historic sites, A.A.Balkema, Rotterdam in print

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lecture McClelland Lecture Analytical contributions to offshore geotechnical engineering Conférence McClelland Contributions des méthodes analytiques à la géotechnique offshore Randolph M. F. Centre for Offshore Foundation Systems, University of Western Australia

ABSTRACT: The theme of this paper, the written version of the 2nd McClelland Lecture, is the contribution of analysis to offshore geotechnical engineering. The application areas considered range from the axial and lateral response of piles, to seabed infrastructure associated with deep water applications, including shallow skirted foundations, anchors, pipelines and risers. The emphasis throughout is on analytical solutions, including appropriately framed outcomes of numerical studies. Most of the material is retrospective, summarising key contributions in an effort to facilitate access, and thus help close the gap between theory and practice. RÉSUMÉ : L’objet de cet article, la 2e conférence McClelland, est de présenter les contributions des méthodes analytiques à la géotechnique offshore. Il couvre plusieurs champs d’application, de la capacité axiale et horizontale des pieux au comportement des structures géotechniques associées aux développements en eaux profondes, incluant notamment les fondations superficielles avec jupe, les systèmes d’ancrages et les pipelines. L’accent est notamment porté sur les solutions analytiques, dont certaines sont basées sur des résultats de solutions numériques. L’essentiel du contenu de cet article résume les contributions antérieures les plus significatives, de façon à en faciliter l’accès et ainsi réduire l’écart entre théorie et pratique. KEYWORDS: Analysis, consolidation, offshore engineering, penetrometers, pile foundations, pipelines, shallow foundations. 1

problem geometry or of the soil response, for example linear elasticity for stiffness solutions, or perfect plasticity for capacity solutions. However, they still provide a framework linking the outcome to the various input parameters, highlighting the critical sensitivities of the response, facilitating parametric studies and quantifying the effect of different idealisations. The paper takes a retrospective look at some of the analytical contributions relevant to offshore geotechnical engineering, drawing attention to the potential application of the solutions in design guidelines and day to day practice. The first part of the paper revisits solutions for the axial and lateral response of pile foundations, which are still the main type of foundation for offshore structures in moderate or shallow water depths and for tension leg platforms in deeper water. The remainder of the paper then focuses more on applications relevant for deep water developments, including subsea foundations, anchors and pipelines. Of necessity, restrictions on the length of the paper have required me to focus on a few specific issues within each topic, in particular where solutions point the way towards improved design recommendations, and recent work addressing developing areas of offshore geotechnical engineering. Before discussing the applications themselves, I should clarify what I intend by the word ‘analytical’ within the present context. I include within this term appropriately conceived parametric studies undertaken through numerical analysis. These should lead to algebraic expressions or charts that may be used in design, identifying the relative contribution of nondimensional groups of parameters that affect the result. By contrast, an algebraic fit through experimental data will rarely provide comparable insight, and should instead be taken as encouragement to quantify the phenomenon through analytical or numerical means. That said, I have always been a strong proponent of the need for high quality experimental data, but with the primary objectives of stimulating understanding of the problem for subsequent analysis, and where necessary to calibrate specific areas of uncertainty in analytical models.

INTRODUCTION

I was privileged to meet Bram McClelland on a few occasions and have always held him in the highest regard. Much of my early exposure to the offshore world was through interactions with the London and Houston branches of the consulting company, McClelland Engineers, that he founded. It was therefore a great honour to be invited to give this, the 2nd, McClelland Lecture, and I am gratified that the written version of the lecture is to form part of the proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering. Following in the footsteps of the first McClelland Lecturer, Don Murff (Murff, 2012), is no easy feat, although I must admit to having become somewhat accustomed to this during my career. More times than I can remember I have found (often retrospectively) that an analytical contribution I have offered has been covered elegantly by Don in a prior publication. It is fitting, therefore, to continue the theme of his own McClelland lecture, in targeting the gap between theory and practice, drawing attention to and summarising various analytical contributions. In an era where virtually any geotechnical application can be modelled numerically, with idealisations potentially limited only to those associated with the constitutive response of the soil, it is tempting to wonder whether true analytical solutions still have a role. At the opposite extreme, design guidelines such as API (2011) and ISO (2003, 2007) are inevitably slow to evolve and in many places rely on somewhat dated suggestions, either empirical or quasi-analytical. There is limited incentive to refine them through analysis without clear evidence of lack of conservatism, or the reverse, excessive conservatism. The potential of analysis is its ability to provide a direct, ideally quantitative, link between a required output and the various input parameters for a given application. At a basic level, dimensional analysis should indicate appropriate nondimensional forms for input and output quantities. Analytical solutions will typically contain idealisations, either of the

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stress is expressed as a function of h/D. These two approaches result in quite similar forms of expression for the shaft friction, but the underlying conceptual models differ. Friction degradation according to h/D, rather than h/Deq seems more logical, since the soil at depths shallower than the pile tip no longer has any knowledge of (or influence from) the area ratio in respect of subsequent densification within the shearing zone adjacent to the pile. The influence of the area ratio on the initial radial stress is also supported by analysis (White et al. 2005). It is acknowledged that the use of the distance, h, to quantify friction degradation is really a surrogate for the number of shear stress cycles to which the soil is exposed, since it is the cyclic shearing that provides the underlying mechanism (White and Lehane 2004). Normalisation by D pre-supposes that piles of different diameter are subjected to broadly similar numbers of hammer blows per diameter advance. Relatively easy or hard driving will affect the rate of friction degradation with h/D. Indeed, ad hoc experimental evidence suggests that hard driving, with limited advance per blow, can cause greater harm due to friction degradation than any benefit obtained by advancing the pile tip further. A missing element from current friction degradation models is any quantified minimum value of shaft friction, below which degradation ceases, because the density of the sand at the pilesoil interface has reached its maximum value for the particular effective stress level. This type of stabilisation has been explored through constant normal stiffness (CNS) shear box testing, and the framework of a predictive model proposed, based on concepts of critical state soil mechanics (DeJong et al. 2006). The secondary influence on the rate of degradation of the cavity stiffness, which is proportional to Gmax/D, would probably be too elusive to extract from the database of pile load tests, but offers a suitable basis with which to refine predictive approaches.

PILE FOUNDATIONS Axial shaft friction

Arguably the most important aspect of pile design, estimation of the profile of limiting shaft friction, has proved resistant to analytical treatment, although understanding of the processes involved has gradually developed. This has allowed appropriate non-dimensional quantities on which the limiting shaft friction depends to be identified. A full discussion of the current design recommendations for shaft friction was provided recently by Jeanjean (2012), and so the remarks below are limited to relatively high level principles underlying the guidelines. In clays and other fine-grained soils, where installation of driven piles occurs over a shorter time scale than dissipation of excess pore pressures, the main quantities to be considered are the undrained shear strength, su, of the sediments, the vertical effective stress, 'v0, and pile geometry: diameter, D, and embedment length, L. It may also be necessary to consider the distance, h, of the element in question from the pile tip. With these parameters as input, empirical correlations have then been used to establish guidelines for the limiting shaft friction, f, normalised by su or 'v0, as a function of su/'v0, L/D and h/D. Other quantities such as the internal angle of friction, and in situ stress ratio, K0, are captured to some extent by the strength ratio, su/'v0, at least within the accuracy of the empirical database. In some clays it may also be necessary to consider the extent to which shaft friction may be limited by a low interface friction angle between pile and soil, or immediately adjacent to the pile, due to the formation of residual surfaces in the clay. For sands, the cone resistance, qc (more strictly the net resistance, qnet) essentially replaces the undrained shear strength in terms of providing a normalising quantity for f and 'v0. The interface friction angle must also be considered, although spanning a relatively small range for typical pile surfaces. The area ratio of open-ended driven piles, relating the crosssectional area of steel to the gross cross-sectional area of the pile, affects the external soil displacement and hence the stress changes in the soil around the pile. For fine-grained soils this will influence the extent of the excess pore pressure field generated during pile installation, and hence the time scale of excess pore pressure dissipation and increase in shaft friction (Randolph 2003), as discussed further below.  It has always been intriguing that the database of pile load tests in clay does not show discernible differences in shaft capacity depending on whether the pile was open-ended or closed-ended (including solid), even though the external stress changes during installation must be affected to some degree. However, cavity expansion analysis shows that, for typical wall thickness ratios (or ratios of Deq/D), the expansion stress is not significantly less (perhaps 15 to 20 %) than for a solid pile, and also some proportion of the total stress increase is lost during the consolidation process, moderating the difference. By contrast, suction caissons have much higher D/t ratios, and even more so when allowance is made for some of the soil displaced by the tip entering the caisson. Hence the final shaft friction will be lower than for a driven pile in similar soil (Randolph 2003). For sands, the area ratio, Ar (or more precisely the effective area ratio, Lehane et al. 2005) influences the magnitude of the radial stresses established in the soil as the pile tip passes, and which subsequently decrease as the pile is driven deeper. A subtle difference among the different cone-based design methods is the manner in which the area ratio is implemented in the estimation of shaft friction (Schneider et al. 2008). In the Imperial College method (Jardine et al. 2005), the shaft friction is taken to degrade from its initial value as a function of the distance, h, normalised by the equivalent diameter, Deq, (where Deq2 = ArD2). By contrast, in the UWA approach (Lehane et al. 2005), while the area ratio is used to modify the ratio of radial stress (close to the pile tip) to qc, the subsequent decay in radial

2.2

Post-installation consolidation

The increase in pile shaft capacity following installation is amenable to analysis, since it corresponds to dissipation of excess pore pressure through (primarily) radial consolidation. Analytical solutions for radial consolidation, following insertion of a solid object such as a pile or piezocone, give the normalised excess pore pressure, U = u/uinitial, as a function of a nondimensional time T = cvt/D2, where cv is the consolidation coefficient (Randolph and Wroth 1979). The solution depends on the rigidity index, G/su, associated with cavity expansion (i.e. the penetration phase). For G/su ~ 100, the relationship between U and T may be approximated by

U

1 1T/ T500.75

(1)

where T50 is the time for 50 % dissipation and is about 0.6. The corresponding value of T90 is about 12. The consolidation coefficient is that associated with radial consolidation and, just as for piezocone dissipation, is biased more towards conditions of swelling, which occurs in the mid to far field, rather than the compression and loss of water content that occurs close to the pile. For an open-ended pile or caisson, the outer diameter, D, should be replaced by the equivalent diameter, Deq, so that T is defined as (Randolph 2003)

cvt cvt T  2 Deq ArD2

(2)

There is very limited field data with which to compare the solution for excess pore pressure dissipation, although some recent studies have reported increases in pile driving resistance

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1 Relative increase in shaft resistance

(Dutt and Ehlers 2009), and of suction caisson extraction resistance (Colliat and Colliard 2010). Figure 1 shows a comparison of the radial consolidation solution with the driving resistance data from Dutt and Ehlers, taken from sites off the coast of West Africa and in the Gulf of Mexico. The long term driving resistance was estimated directly from the API design guidelines, since the longest re-drive delay was only 8 days (West Africa) to 12 days (Gulf of Mexico). The data were plotted together, even though the pile diameters varied between 2.7 m (West Africa, diameter to wall thickness D/t = 40, so Deq = 0.85 m) and 2.1 m (Gulf of Mexico: D/t = 48, so Deq = 0.6 m). The initial driving resistance was around 20 % of the (estimated) long term resistance, so the analytical consolidation solution has been adjusted to give a proportion of long term resistance of 0.2 + 0.8U. The solution matches the Gulf of Mexico data reasonably, with a plausible consolidation coefficient of cv = 20 m2/yr. The data from West Africa do not show a clear trend, but are mostly bounded by a theoretical curve for cv = 100 m2/yr. Although this seems quite high, these piles were driven to a depth of 150 m, twice the depth of the Gulf of Mexico piles, and so is reasonable as an upper bound. Data from suction caissons from offshore West Africa are shown in Figure 2. The suction caissons were extracted (by pumping water into them) at different periods following installation (Colliat and Colliard 2010). The diameters ranged between 3.8 and 8 m, and penetration depths from 16.5 to 20.5 m. Although much greater diameter than typical driven piles, the values of wall thickness were only 20 or 25 mm. Allowing for only 50 % of the soil displaced being pushed outwards (Zhou and Randolph 2006), the equivalent diameters are only 0.28 to 0.45 m. The relative increase in shaft resistance has been obtained by normalising by the original shaft resistance. The longest elapsed time was 1260 days, where the reported shaft resistance was 2.03 times the installation value (the data point is plotted at a reduced time of 100 days, in order to limit the time axis). All data points on Figure 2 have been plotted after first scaling the actual time by (0.3/Deq)2 in order to give a common basis of comparison. Inevitably there is some scatter in the data, but the theoretical consolidation curve for cv = 10 m2/yr (and Deq = 0.3 m) lies within a factor of about 2 for all but one datapoint. The coefficient of consolidation seems reasonable, given that the average depth is almost an order of magnitude lower than for the driven piles in Figure 1.

0.9 0.8

Radial consolidation solution (cv = 10 m2/yr; Deq = 0.3 m)

0.7 0.6 0.5 0.4 0.3

Data from suction anchors (Colliat & Colliard 2011)

0.2 0.1 0 0.1

1 10 Time (days) - scaled for Deq = 0.3 m

100

Figure 2 Increase in suction caisson extraction resistance with time following installation.

The time scale of consolidation reported by Colliat and Colliard (2010) is similar to that noted by Jeanjean (2006), for suction caissons with diameters 2.9 to 3.7 m (equivalent diameters of 0.39 to 0.53 m). Unfortunately, though, the latter dataset did not include any short term restart or retrieval data, with the earliest being after a time delay of 50 days (equivalent to 16 days for Deq = 0.3 m). As such, all cases showed relative increases in excess of 50 %. The average long term (~1000 day) increase in shaft resistance was only 75 %, compared with 100 % for the West Africa suction caisson data. It is perhaps disappointing that greater use is not made of rigorous consolidation analysis in estimating the time scale for the increase in shaft resistance of piles and suction caissons. Commentary on the topic is partly obscured by musings on thixotropy, which may play a role but with no guidance provided on how to scale from laboratory to field. Ultimately the shaft resistance results from the increase in normal effective stress, which is adequately modelled by consolidation analysis. 2.3

Axial load-displacement response

In the offshore industry it is customary to use load transfer methods to evaluate the axial load-displacement response. Nonlinear load transfer curves allow the full pile response to be evaluated, from the initial quasi-linear response right up to failure. It is instructive, though, to consider the form of the load transfer curves, and elastic solutions for the complete pile that are applicable at low load levels. Analytical solutions for axial pile response abound, with gradually increasing degree of sophistication, starting with Murff (1975) for the case of a linear load transfer stiffness, ka, uniform with depth. Randolph and Wroth (1978) related the load transfer stiffness to the soil shear modulus, G, and extended the solution in an approximate manner to consider a linear variation of modulus with depth. This was later extended in a more rigorous manner by Guo and Randolph (1997) for power law variations of modulus with depth, and by Mylonakis and Gazetas (1998) for layered profiles, and with allowance for interaction effects between piles. The solutions for uniform soil modulus with depth may be expressed in the generic form of

cv = 100 m2/yr; Deff = 0.85 m

Radial consolidation solution (cv = 20 m2/yr; Deq = 0.6 m)

Pt K  S tanhL  K axial   S b wt S  K b tanhL 

(3)

with Figure 1 Increase in pile shaft capacity with time following driving (field data and original figure from Dutt and Ehlers 2009).

k L  a L and S EA p

87

EAp L

 L

EAp k a

(4)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

where Pt and wt are the load and displacement at the top of the pile, Kb is the base stiffness (Pb/wb), L the embedded pile length and (EA)p the cross-sectional rigidity of the pile. The solution may be extended for linear variation of modulus with depth by pre-multiplying the tanh(L) term in the numerator by , the ratio of average modulus to that just above the pile base (Randolph and Wroth 1978); for layered profiles, the base stiffness, Kb, can be replaced by the load-displacement stiffness of the pile segment below the one under consideration, nesting subsequent layers in the same way. The load transfer stiffness, ka, (ratio of axial load transfer per unit length of pile to the local axial displacement) may be related to the soil shear modulus, G, by

ka 

2  2L  G where  ~ ln ~4   D

The analytical solution for the pile head stiffness allows the effect of pile compression (or extension), which is controlled by the quantity L, to be explored. For a stiff pile (high ratio of (EA)p/L to kaL), the overall pile head stiffness, Kaxial, is just the sum of the shaft and base stiffness acting in parallel (i.e. Kb + kaL). However, as L increases, tanh(L) approaches unity and the pile head stiffness asymptotes to

EAp k a ~ 1.25 EAp G K axial  S

The above relationship is useful for estimating the dynamic stiffness of a pile (substituting G0 for G). It also provides a guide to evaluate the load at which failure first occurs at the pile-soil interface, which may be expressed as

(5) Pslip

Randolph and Wroth (1978) provided more explicit guidance on the parameter , which arises due to a logarithmic singularity in integrating the shear strains around the pile. However, within the accuracy to which G may be determined, a value of 4 is sufficiently accurate for piles of moderate L/D. The ratio of shear strain in the soil adjacent to the pile to the normalised displacement, w/D, is given by /2 (i.e. about 2). This leads to a first estimate for the pile displacement required to mobilise full shaft friction as wf/D ~ 2f/G (where f is the limiting shaft friction), which would fall in the range 0.5 to 2 % for G/f of 100 to 400. For a hyperbolic soil response where the secant shear modulus decreases inversely with the strength mobilisation, /f, the parameter  may be replaced by (Kraft et al. 1981)

 ~ 4  ln(1  ) where   Rf

 f

Qshaft

or

k a initial ~ 1.5G 0



1 1  L L

EAp ka

~

0.8 L

EAp

(9)

G

This has particular relevance for assessing the cyclic robustness of piles under axial loading. There is substantial experimental evidence that suggests degradation in load transfer under cyclic loading occurs very rapidly once local slip has occurred (Erbrich et al. 2010). Stability diagrams for cyclic loading are generally expressed in terms of the cyclic and mean loads applied at the pile head, normalised by the pile (shaft) capacity, as illustrated in Figure 3 (Poulos 1988, Puech et al. 2013). However, such diagrams do not take account of the relative compressibility (or extensibility) of the pile within the soil. For high ratios of (EA)p/GL2, slip will occur at relatively low proportions of the shaft capacity, which will allow degradation to occur, reducing the shaft friction in the upper part of the pile to a cyclic residual level.

(6)

with the hyperbolic parameter, Rf, typically around 0.9 to 0.95. This gives a reduction in secant load transfer stiffness by a factor of approximately 2 between low and high shaft friction mobilisation. More general forms of hyperbolic soil model, such as suggested by Fahey and Carter (1993), may be integrated to provide alternative estimates for the evolution of the load transfer stiffness.  The generic form of axial load transfer curves suggested in the offshore guidelines are consistent with this reduction in secant stiffness, with normalised ratios of (/f)/(w/wf) that reduce from 1.875 to unity. In a welcome step forward, the latest version of the API guidelines (API 2011) now recommends a similar shape of load transfer curve, and mobilisation displacement, wf, for sand as for clay, replacing the previous recommendation of 2.5 mm for sand (an anachronism based on experimental data for relatively small pile diameters). Jeanjean et al. (2010) outlined the logic for mobilisation distances for sand, with correlations for G/'v0 and /'v0 suggesting values around 0.5 % of the diameter, but experimental data generally grouped above 1 % of the diameter. The net result was to propose a similar range for the displacement, wf, to mobilise failure, for both sand and clay, in the range 0.5 to 2 %. The underlying theoretical link between the load transfer stiffness and the soil shear modulus should, however, be borne in mind. Where values of small strain shear modulus are available, it would be more sound, theoretically (particularly for assessing dynamic stiffness), to link the initial load transfer gradient to the small strain shear modulus of the soil. Thus the initial gradient should be G  d  ~ 0   dw   initial 2 D

(8)

1

Normalised cyclic load, Qcyclic/Qshaft

0.8 Unstable N < 10

0.6

Increasing cycles (N) to failure

N ~ 300

0.4

Metastable 0.2

Stable N > 10,000

0 0

0.2

0.4

0.6

0.8

1

Mean load, Qmean/Qshaft Figure 3 Typical form of cyclic stability diagram.

Cyclic stability diagrams are therefore of limited use for a complete pile (unless it is relatively stiff), although they are useful to describe the soil response at a local level, rather like similar diagrams for element tests (Andersen 2009). An alternative approach is to use shakedown theory to arrive iteratively at a profile of mean and cyclic shear stresses down the pile that all lie within the stable zone of a stability diagram (based on soil element response). Residual shaft friction conditions should first be assumed throughout the upper region of the pile where slip occurs under the maximum operational loading.

(7)

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2.4

With this definition, the reduction factor can be approximated as

Post-peak strain softening

Axial compression or extension of the pile leads to non-uniform mobilisation of shaft friction down the length of the pile, with slip between pile and soil generally being initiated at the mudline and gradually propagating down the length of the shaft. Any strain softening in the load transfer response will therefore allow a form of progressive failure, such that the maximum shaft resistance will be less than the ideal value for a hypothetical rigid pile. Alternative forms of load transfer curve are shown in normalised form in Figure 4, with the linear degradation to 70 % of peak shaft friction being consistent with API and ISO design guidelines for clay. A difficult consideration is how to scale the degradation response from laboratory to field scale, and the extent to which a given degree of degradation should be based on absolute displacement or displacement scaled to pile diameter. Even though the degradation occurs locally at the pile-soil interface, the surrounding stress field (and even the width of the main shearing zone) is affected by the pile size, so that scaling with pile diameter seems at least partly justified. In some soil types, much more significant degradation can occur (Erbrich et al. 2010), possibly occurring over rather greater displacement than the 1 % of pile diameter suggested in offshore design guidelines.

 1   R pf ~   1   tanh   C

as illustrated in Figure 5. This expression provides an initial estimate to assess the extent to which progressive failure may reduce the net shaft resistance. The actual reduction factor will depend on the precise form of the load transfer curve, particularly in respect of degradation, and should therefore be evaluated through numerical analysis. A detailed investigation of the performance of steel jacket structures in Gulf of Mexico hurricanes found that the one case where axial pile failure occurred could be explained by progressive failure using the API (2011) recommended form of load transfer curve with degradation to 70 % of peak friction (Gilbert et al. 2010). 1

Reduction factor, Rpf

 = 0.75

1

Normalised shear stress, /f

linear exponential

fully brittle

0.75

wres/wf

 

C

D f L2 EAp w f  w res 

0.5

 1   R pf ~   1   tanh   C

 = 0.25 0.25

0

2

2.5

3

Normalised displacement, w/wf

Figure 4 Alternative forms of post-peak softening in axial load transfer curves.

The actual shaft resistance, Qs, may be expressed as a proportion, Rpf, of the ideal shaft capacity, Qshaft:

Qs  R pf Qshaft where Qshaft  DLf

(10)

The value of Rpf will be a function of the degree and brittleness of strain softening and the compressibility of the pile. An analytical solution for the extreme case of ‘ìnstant’ strain softening was given by Murff (1980), who expressed the reduction factor, Rpf, as a function of the strain-softening ratio,  = res/f, and a non-dimensional pile compressibility, 3. The latter quantity may be shown to be identical to L. For strain softening over a finite distance, wres, Randolph (1983) proposed an alternative non-dimensional pile compressibility or compliance, C, substituting the displacement to failure, wf, for the displacement from peak to residual, wres. Numerical experiments suggest, however, that a more robust measure of pile compliance, in respect of progressive failure, is the total displacement to residual, i.e. wf + wres, with C defined as:

D f L2 EAp w f  w res 

1.5

2

2.5

3

Lateral pile resistance – clay

Design methodology for the lateral response of piles is almost universally based on load transfer approaches. These are wellsuited to capture the significantly non-linear soil response, particularly in the upper few diameters of the pile. However, the proposed load transfer curves are labyrinthine in formulation and with no obvious link to any analytical basis. Jeanjean (2009) has argued for an overhaul of the API guidelines for soft clay conditions, proposing an alternative formulation based on a combination of (centrifuge) model test data and finite element analysis, but with the ultimate lateral resistance at any depth linked to upper bound solutions (Murff and Hamilton 1993). The Murff and Hamilton solution addresses soil failure at shallow depth, based on a three-dimensional conical wedge mechanism. Below the wedge, the lateral resistance is limited by plane strain flow around the cylindrical pile (Randolph and Houlsby 1984, Martin and Randolph 2006). The solutions take account of the relative roughness between pile and soil, with the limiting (plane strain) resistance at depth varying with the friction ratio,  as:

0 1.5

1

Inverse of square root of pile compliance, C

2.5

1

0.5

Figure 5 Reduction factor due to progressive failure.

0.25

C

 = 0.5

0



0.5

0.75

=0

0.5

0

(12)

Pu  N p ~ 9.14  4.14  1.34 2 su D

(13)

From a design perspective, a simple linear fit of Np = 9 + 3 is sufficiently accurate, being generally about 3 % conservative apart from at the limit of a fully rough pile when it rounds to 12 instead of 11.94. There is an incompatibility at the transition depth between the wedge and the plane strain flow, but this does not appear to have a significant effect on the overall pile resistance, judging by comparisons with full finite element analyses. The discontinuity can be removed by allowing a gradual transfer

(11)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Although Jeanjean’s study was for soft clays, in principle the same general approach should be applicable to stiff clays but with some caveats:  Where stiff clays occur at the seabed surface, a gap is much more likely to develop than for soft clays (since higher su/'D, and much greater suctions required to be sustained in order to prevent a gap forming). As such, the surface factor (N1 - N2) should be halved, while retaining the same limiting (plane strain) value of Np.  A lower friction ratio, , is likely to be appropriate, just as for axial shaft friction.

from the wedge mechanism to the flow mechanism (Klar and Randolph 2008). Although the Murff and Hamilton upper bound solution treats the conical wedge mechanism as a whole, to provide an overall lateral resistance for that section of the pile, they explored suitable variations of Np with depth, z, that fitted the overall upper bound resistance for piles of different embedment. This led to proposed factors of

 N p N1  N 2e  z / D

(14)

with  adjusted for different strength profiles idealised as su = sum + z, according to    s  Min 0.25  0.05 um , 0.55 D  

2.6

Lateral pile resistance – sand

For sand, design recommendations for limiting lateral resistance still rely on a limit equilibrium calculation for a putative passive wedge of soil failing ahead of the pile. There is also an overriding maximum limiting resistance, proportional to depth, although this is extremely high (such that, in practice, it would not be reached shallower than depths exceeding 15 pile diameters). The resulting profiles of limiting resistance are not consistent with results from numerical modelling, or even with empirical data that appear to follow a linear trend, below a depth of about 1 diameter, that is broadly proportional to the square of the passive earth pressure coefficient, Kp. However, any design approach requiring what is ultimately a bearing resistance, but is couched in terms of friction angle, ', suffers from the problems of (a) how to ‘measure’ ', and (b) the need to adjust ' according to the resulting implied effective stress level. Typically values of ' must be deduced from the results of cone penetration tests. It is therefore far more logical to link the lateral pile resistance directly to the cone resistance, following the path taken for axial pile capacity. Empirically based approaches that express the lateral pile resistance as a function of the cone resistance have been proposed for carbonate sands (Wesselink et al. 1988, Novello 1999, Dyson and Randolph 2001). Recently, a numerical study has been undertaken by Suryasentana and Lehane (2013) to provide a more theoretical link between lateral pile resistance and cone resistance, the latter being simulated as spherical cavity expansion. Material properties were based on those for a typical silica sand. Systematic dimensional analysis, with a parametric study covering a wide range of the various dimensionless groups, allowed relationships to be developed between normalised values of pile resistance, cone resistance, depth and lateral displacement. The eventual relationship incorporated an exponential term to give a true limiting lateral resistance at large displacement. The lateral resistance was then expressed as (Suryasentana and Lehane 2013):

(15)

The value of Np therefore increases from a surface value of N1 – N2, to a limiting value at depth of N1 (corresponding to Equation (13)). Assuming a double sided mechanism (with negative excess pore pressures behind the pile causing the soil to move with the pile) the Murff and Hamilton mechanism leads to an almost constant value of 5 for N2. Thus the surface value of Np increases approximately linearly with  from about 4 for a smooth pile ( = 0) to 7 for a rough pile ( = 1). Jeanjean (2009) has recommended adoption of N1 = 12 and N2 = 4, without consideration of the friction ratio, . Even for fully rough conditions this is slightly optimistic in respect of the surface value of Np (8 instead of the upper bound value of 7). Also, as commented by Murff and Hamilton (1993), the additional resistance provided by a fully rough pile compared with a smooth pile “would seem to be particularly susceptible to degradation due to cyclic loading, and thus it may not be prudent to count on it for design”. A compensating factor to this (intuitive) consideration is the gradual hardening that occurs due to consolidation between periods of cyclic perturbation (Zhang et al. 2011). The net effect of this is that the post-cyclic monotonic pile responses showed slight increases in resistance for a given pile displacement. Similar hardening was observed in centrifuge model tests simulating the interaction of steel catenary risers with the seabed (Hodder et al. 2013). Equally important for lateral pile design is the mobilisation of lateral resistance with displacement. Variations in the stiffness at small displacements for elements at some depth down the pile can have a significant effect on the pile head response. The current API and ISO guidelines for load transfer curves appear too soft at moderate displacements (Jeanjean 2009), although the initial data point, with P/Pu = 0.23 for a displacement of y = 0.1yc = 0.2550D, implies a rather high stiffness. Here 50 is defined as the strain in a (triaxial) compression test at half the failure deviator stress, which is equivalent to su/3G50. Hence for Pu = 9suD, the initial gradient is P/y = 9×0.23×3G50/0.25 = 25G50. Theoretical solutions for the load transfer response, either based on an analogy with cavity expansion or closed form solutions (Baguelin et al. 1977), lead to a gradient of kpy ~ 4G, and hence a maximum gradient of 4G0. Applying this as a limit at small displacements to the hyperbolic tangent function suggested by Jeanjean (2009) leads to    G P y  4G 0  ,  Min  tanh 0.01 0 y (16)  Pu s u D  Pu    

 q   2 c  v 0 D  v 0  P

0.68

z    D

0.61 

1.1 0.94   y  1  exp  8.9 z        D D   

(17)

This study represents an important step towards a more rational approach to the estimation of load transfer responses for lateral pile design in sand. The rather gradual development of the ultimate resistance (the terms outside the square bracket in Equation (17)) is in stark contrast to the hyperbolic tangent relationship in the current design guidelines, which leads to the ultimate resistance being mobilised at displacements of 1 or 2 % of the pile diameter. 3

SHALLOW FOUNDATIONS

Design guidelines for shallow foundations that are provided in the main geotechnical guides (ISO 2003, API 2011) have developed from guidance for temporary mudmat foundations to support steel jacket structures, prior to pile installation. Large

For Pu = 12suD, the transition point occurs at y/D = 0.0009, so P/Pu = 0.0003G0/su or 0.12 for G0/su = 400.

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expressions for failure envelopes. As is the nature of analysis, idealisations of the real system have to be made, with each study tending to focus on a different set of restrictions. There is a wealth of information in the various contributions, to which justice cannot be done here. Instead, one or two salient points will be commented on and suggestions made for practical approaches for use in design. The focus is on rectangular foundations, with relatively shallow skirts, since these are of particular relevance to deep water developments. A schematic of the problem is shown in Figure 6. In the most general case, six independent loads and moments may act on the foundation, and the dimensionless groups that need to be considered include the foundation aspect ratio, B/L, embedment ratio, d/B, and normalised soil strength gradient,  = B/su0. (Relevant ratios for a circular foundation of diameter, D, where the loading can generally be simplified to the three components, V, H and M, in the plane containing the resultant horizontal load, are d/D and D/su0.)

gravity foundations and spudcan foundations are dealt with in separate documents focusing respectively on concrete structures and mobile drilling rigs. The main geotechnical guidelines focus on bearing capacity, based on classical solutions for strip foundations, modified using heuristic adjustment factors for foundation shape and embedment, and the influence of horizontal and moment loading. The largest use of shallow foundations offshore is now for deep water subsea systems, where they are widely used for pipeline end terminations and manifolds. In the main, the seabed sediments in deep water comprise fine-grained soil, with relatively low strength at mudline. The foundations are steel mats, generally rectangular in plan with shallow skirts, and with a high cost incentive to minimise the size to allow installation from pipe-lay vessels. The emphasis in design for geotechnical capacity is on horizontal and moment loading from the attached pipeline and jumpers, rather than on vertical bearing capacity. Along with the changing nature of shallow foundation applications, the last decade or so has seen increasing analytical emphasis on the development of failure envelopes in vertical (V), horizontal (H) and moment (M) load space. The most recent API guidelines (API 2011) now include a commentary that permits (encourages would be too strong a word) the use of failure envelopes as an alternative approach; this is timely since it suits better application to shallow foundations for subsea systems, where failure tends to be by sliding or overturning.

L Mx

B mudline

LRP

Max B/su0 or D/su0

Max embed. depth d/B

Strip, Circle, Rectangle

Tension, Closed form (*)

∞

0

S

Y*

Bransby-Randolph 1999

VHM

6

0.17

S

Y*

Houlsby-Puzrin 1999

VHM

0

0

S

N*

Taiebat-Carter 2000

VHM

0

0

C

Y*

Taiebat-Carter 2002

VM

0

0

C

N

Gourvenec-Randolph 2003

HM

10

0

SC

Y

Randolph-Puzrin 2003

VHM

6

0

C

Y*

Load cases VHM

Finnie-Morgan 2004

HT

0

0

SCR

-

Yun-Bransby 2007

HM

200

1

S

Y

Gourvenec 2007a

VHM

0

0

R

N*/Y

Gourvenec 2007b

VHM

6

0

SC

Y

Gourvenec 2008

VHM

0

1

S

Y

VHM

200

1

S

Y

Yun et al. 2009

VHT

0

0

SCR

-

Taiebat-Carter 2010

VHM

0

0

C

N

sum su0

su 

z

z

Figure 6 General loading applied on a rectangular skirted foundation with linearly varying soil strength.

Even though typical embedment ratios of subsea system foundations are quite low, there can still be an appreciable increase in capacity. Design guidelines simplify the effects of aspect ratio and embedment into separable additive factors, whereas in reality the depth factor is itself a function of the embedment ratio (Salgado et al. 2004) and also the strength gradient factor, . The depth correction factor in API (2011) is deliberately conservative (Figure 7), expressed as: d c  1  0.3 arctan d / B

(18)

with B replaced by the effective width, B', for foundations where no tensile stresses are permitted. 1.6 Depth factor 1.5 dc 1.4

Circle (Martin 2001) D/su0 = 0 Inverted parabola

1

1.3 Rectangle (B/L = 0.5)

1.2

2

Salgado et al. (2004) (strip foundation) 1+0.3arctan(d/B) 5 and 10 (circle)

1.1

Bransby-Yun 2009

yd

T V

Reference

Hy My

Table 1 Summary of analytical and numerical studies of failure envelopes for shallow foundations for undrained conditions

Bransby-Randolph 1998

x Hx

1 0

0.2

0.4

0.6

0.8

1

Embedment, d/B or d/D

Murff et al. 2010

HT

0

0.05

R

-

Gourvenec-Barnett 2011

VHM

6

1

S

Y

Feng et al. 2013

Full 3D

10

0.2

R

Y*

Figure 7 Depth correction factors for different shaped foundations.

This expression provides a lower bound to those derived analytically, even for a strip foundation. The correction factor from Salgado et al. (2004) for strip foundations, which varies with the square root of d/B, is shown in Figure 7 for comparison. Also plotted are depth factors deduced from lower bound results for circular foundations for a range of D/su0

Table 1 provides a summary of some of the solutions published over the last fifteen years, indicating which include either analytical solutions, or at least closed form algebraic

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

(Martin 2001). These show the effect of D/su0 (), with the depth factor reducing with increasing  as d/D increases, relative to the factor for homogeneous (uniform strength) soil. In the range relevant for subsea systems, the results for different values of  converge, and can be fitted by an inverted parabola with apex at dc = 1.5 for d/D = 2. However, these are still lower than for a rectangular foundation with B/L = 0.5, according to results of 3D finite element analyses (Feng et al. 2013). These give an initial gradient for the depth factor of greater than unity with respect to d/B, and a significant 17 % increase in bearing capacity for d/B = 0.2. For circular foundations, it is possible to develop threedimensional failure envelopes in V-H-M space. Failure envelopes are most effectively expressed in normalised units, v = V/Vu, h = H/Hu and m = M/Mu, where the subscript “u” indicates the limiting uniaxial resistance (e.g. for V, with M = H = 0). A promising form for foundations that can withstand tensile stresses is (Taiebat and Carter 2000):

the H and M planes. These maximum values are then reduced further according to the mobilised torsion, t, by considering h-t and m-t interaction diagrams. The logic behind the various steps is to arrive at a final h-m failure envelope that has already taken full account of the mobilisation ratios for vertical and torsional modes of failure. Full details of these steps are described by Feng et al. (2013) for rectangular skirted foundations that can withstand tensile stresses. The failure envelopes involving v are based on generic shapes proposed in the literature, for example v v *  1  v * 1  h q for v  v * else h  1 

with v-t interaction following a similar type of envelope as for v-h interaction. Values of the transition v (v*) and the exponents q, p have been fine-tuned for rectangular foundations with B/L in the region of 0.5, and take account of the loading direction relative to the rectangular foundation and (for p) the normalised shear strength gradient. Other failure envelopes, for hx-hy, h-t etc are elliptical in form, for example

2

  m  v   m1  0.3h   h 3  1 0  m     2

(19)

which gave a reasonable fit to finite element results for a circular foundation resting on the surface of homogenous soil. An improved failure envelope, though not expressed in algebraic form, was discussed by Taiebat and Carter (2010). The various powers and coefficients would need adjusting for different foundation shapes, embedment ratios and normalised shear strength gradient. There is little prospect of any simple way of expressing a failure envelope for full three-dimensional loading applied to a rectangular foundation. Instead, a simplified approach has been proposed recently (Feng et al. 2013), taking advantage of the relatively low mobilisation of the uniaxial vertical capacity for subsea system foundations, where unfactored values of v will rarely exceed about 0.3.

h ax  h by  1

q

     m d m d  1  h m d  h 2   h 2  1 0     md   md  

2

Evaluate uniaxial capacities for vertical, horizontal, moment and torsional loading

3

Reduce ultimate horizontal, moment and torsional capacities to maximum values available, according to mobilised (design) vertical capacity, v = V/Vu

4

Normalised moment, m = Md/Mu (θm = 30°)

For given foundation geometry evaluate su0 and nondimensional quantities B/L, d/B and 

For given angle, , of resultant horizontal load, H, in the horizontal plane, evaluate corresponding ultimate horizontal capacity, and similarly for ultimate moment capacity

5

Evaluate reduced ultimate horizontal and moment capacities due to normalised torsional loading

6

Evaluate extent to which applied (design) loading falls within H-M failure envelope, and thus safety factors on self-weight V, live loading H, M, T or material strength su0

(22)

where the parameters q,  and  are expressed as functions of  and, in the case of  as a function of the resultant horizontal loading direction,  = arctan(Hx/Hy) (Feng et al. 2013). It was found that the shape of the failure envelope became insensitive to the embedment depth provided the moment was expressed as if the load reference point was shifted from mudline to skirt tip depth, d; thus Md = M + Hd. 

Details

1

(21)

again with each envelope fitted to results from 3D finite element analyses, expressing the exponents a and b as functions of the dimensionless input variables. The final form of h-m failure envelope is similar in nature to that proposed by Taiebat and Carter (2000), although now without the term for v (which has been allowed for separately):

Table 2 Steps in design process for subsea system foundations Step

(20)

 v 1  m p

The steps in the approach are tabulated in Table 2. In common with most failure envelopes, the uniaxial capacities are first evaluated, providing a first indication of the relative mobilisation for each of the 6 degrees of freedom. Using interaction diagrams for v-hx, v-hy, v-mx, v-my and v-t, reduced allowable values of Hx, Hy etc are deduced, according to the applied v. Separate interaction diagrams for hx-hy and mx-my (with the ultimate values for each component reflecting the reduction from the previous step) then allow estimates of the maximum resultant H and M, for the given loading angles in

1.2 T/Tu= 0, 0.25, 0.5, 0.75, 0.9 1 0.8 0.6 0.4 0.2 0 -1

-0.8

-0.6

-0.4

-0.2

0

0.2

0.4

0.6

0.8

1

Normalised horizontal load, h = H/Hu (θ = 60°) FE results

Estimation

Figure 8 Example comparison between estimated failure envelopes for different torsion mobilisation ratios and FE results (Feng et al. 2013).

Examples of the fit between results of individual finite element computations and the estimated failure envelopes are

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capacity as well. For rectangular foundations all six degrees of freedom need to be considered. Generic shapes of failure envelope, based on loads normalised by their ultimate uniaxial values, are much less sensitive to foundation shape and embedment ratio, and soil strength gradient, than are the uniaxial load limits. As such, the shapes need not necessarily be fine-tuned. The most awkward shape is the failure envelope in the h-m plane. For planar loading, the approach described by Gourvenec (2007b) is therefore attractive, based on generic failure envelopes in v-m space for different magnitudes of (normalised) horizontal load eccentricity, m/h. For rectangular foundations, particularly if relatively lightly loaded vertically, the approach outlined in Table 2 offers a simple way forward, maintaining a modular concept where the various interaction diagrams may be fine-tuned to suit particular conditions, if these deviate significantly from those considered by Feng et al. (2013). For example, interaction diagrams based on sustained tensile stresses could be replaced by equivalent ones based on a zero tension condition. The increasing complexity of subsea systems brings the potential for higher service loads due to thermal and pressuredriven movements of the pipeline and jumper connections. The cost incentive to limit the overall foundation dimensions is therefore driving innovation, both in analysis methods but also in the foundation configuration itself. One such innovation is to include pin-piles at the foundation corners, which can increase the sliding and torsional capacity by a factor of 3 or 4. A simple design approach for such a hybrid foundation has recently been developed, following lower bound principles (Dimmock et al. 2013), and validated through physical model tests (Gaudin et al. 2012). An alternative approach is to design the foundation to slide, hence reducing the magnitudes of horizontal load and moment (Bretelle and Wallerand 2013). Both of these strategies still rely on failure envelopes for different combinations of load and moment, either to ensure adequate capacity, or to evaluate the displacement and rotation paths for sliding foundations.

shown in Figure 8, for a case of a surface foundation on homogeneous soil, with resultant horizontal loading at 60 º to the x-axis. The failure envelopes and FE results correspond to five different torsion mobilisation ratios. The quality of fit is reasonably good, although with slight over prediction of the maximum moment capacity at high levels of torsion. An example foundation analysis following this approach is presented here, with input data (including factored design loads) tabulated in Table 3 and the resulting failure envelopes and design loading shown in Figure 9. Failure envelopes based on unfactored shear strengths are shown as dashed lines, with the outer (black) envelope corresponding to zero torsion, and the inner (red) envelope after allowing for the applied torsion of 2100 kNm. The solid lines represent failure envelopes after reducing the shear strength by the material factor of 1.58 that is just sufficient to cause failure; again the outer and inner of these two envelopes represent situations with zero torsion and the actual design torsion. The increased mobilisation ratios for v and t, due to factoring the shear strength, reduce the maximum values of H and M for the failure envelopes that allow for the applied torsion by greater factors, respectively 2.1 and 1.8. Table 3 Input data for example subsea system foundation Parameter

Value

Units

Design loads

Value

Units

Width, B

8

m

Vert. load, V

1200

kN

Length, L

16

m

Load, Hx

200

kN

Skirt, d

0.6

m

Load, Hy

300

kN

Strength, sum

5

kPa

Moment, Mx

1500

kNm

su gradient, k

2

kPa/m

Moment, My

-2400

kNm

Skirt friction

0

Torsion, T

2100

kNm

10000

Resultant moment, M (kNm)

V = 1200 kN

9000

4

8000 7000 Zero torque

In most design applications, failure envelopes are used to establish safe load combinations. However, they may also be used to model the kinematic response during continuous failure. The concept was applied to predict the trajectory of drag embedment anchors by Bransby and O’Neill (1999), successfully simulating centrifuge model tests (O’Neill et al. 2003). In soft sediments, drag anchors embed to several times the length of their flukes, advancing approximately parallel to the flukes and gradually rotating until the flukes approach the horizontal, signifying reaching their ultimate penetration depth. The anchor chain forms a reverse catenary through the soil, described by an analytical solution expressed in terms of the chain tension, T, and average soil resistance, Q , between mudline and padeye depth (Neubecker and Randolph 1995). Critical is the angle change between mudline and padeye, which may be approximated as

T = 2100 kNm

6000 5000 4000

Unfactored su

Design point

3000 Factored su

2000 1000 0

-1000

-750

-500

-250

0

250

500

750

USE OF FAILURE ENVELOPES FOR ANCHORS

1000

Resultant horizontal load, H (kN) Figure 9 Failure envelopes and design loading for example application.

From a design perspective, optimising the size of shallow foundations for subsea systems requires more sophisticated analysis than the conventional approach for bearing capacity followed in offshore design guidelines. The use of failure envelopes for combined V-H-M loading provides a suitable advance. Depending on the sensitivity of the structure, final design may well involve detailed 2D or 3D finite element analysis, but simpler tools are needed to enable initial sizing. Design using failure envelopes is modular, with the first step being to evaluate uniaxial failure loads and moments for the relevant degrees of freedom. For circular foundations in-plane loading may generally be assumed, with only three degrees of freedom, unless the torsion is significant. If that is the case, the horizontal capacity should be reduced to compensate (Finnie and Morgan 2004, Murff et al. 2010), and possibly the moment



2 a



 02 ~

2z a Q Ta

(23)

where subscripts ‘a’ and ‘0’ correspond to the anchor padeye and mudline respectively. Solutions for the final anchor embedment depth and ultimate capacity were initially obtained using simplified limit equilibrium (Neubecker and Randolph 1996) or upper bound (Aubeny et al. 2005, 2008) approaches. The use of a full failure envelope to obtain the relative motions, parallel and normal to the anchor fluke, and rotation, represented a more rigorous treatment.

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The relatively large projected areas of 10,000 mm2 for the standard T-bar, and generally 3000 to 5000 mm2 for the ball penetrometers used offshore, makes them attractive for characterising soft clay deposits, but still with a capability to penetrate sand layers with cone resistance of up to 3 or 4 MPa. In particular, full-flow penetrometers have become the de facto standard for strength profiling in the upper few metres, with application to pipeline and riser design. Measurement of remoulded resistance from cyclic tests, which also help to constrain the accuracy of the monotonic penetration data, is essential for pipeline design. While both geometries are used, with the T-bar having superficial similarity to an element of pipe, the ball is a kinder geometry and has the advantage of enabling pore pressure measurement, as discussed later. Plasticity solutions for the T-bar and ball in ideal (nonsoftening, rate independent) soil give resistance factors that may be approximated by Equation (13) or NTbar-ideal = 9 + 3, and

The form of failure envelope adopted by Bransby and O’Neill (1999) was based on that suggested by Murff (1994):



nq  mr  st



1/ p 

 1 0

(24)

where n, m and s represent the mobilisation ratios (e.g. n = N/Nu) for normal, moment and sliding modes relative to the anchor fluke. Ultimate, uniaxial, limits, Nu, Mu and Su are typically obtained from a combination of plasticity solutions and finite element analysis, depending on the anchor fluke shape (O’Neill et al. 2003, Aubeny and Chi 2010). Similarly, the various powers may be adjusted to fit different anchor shapes, with q and t typically in the range 3 to 5, and p, r around unity (Bransby and O’Neill 1999, Elkhatib 2006, Yang et al. 2010). The values of q, r and t should not be chosen less than p, in order to guarantee convexity of the failure envelope. A similar approach was adopted to model the keying of mandrel-installed plate anchors, such as the suction embedded plate anchor or SEPLA (Cassidy et al. 2012, Yang et al. 2012). Combining the chain response with the failure envelope allows the full kinematic response of the plate anchor to be investigated. The position of the padeye relative to the plate centre may then be optimised, minimising loss of embedment during keying or even such as to cause the anchor to dive. A careful finite element based parametric study showed that the original SEPLA design, which incorporated a hinged flap to help limit loss of embedment during keying, was ill conceived (Tian et al. 2013). More recent numerical work has considered sophisticated 3D anchor geometries, investigating how the presence of the shank affects the failure envelope (Wei et al. 2013). 5

Pu 0.25 D 2 s u

 N ball  ideal ~ 11.21  5.04   1.06 2

(25)

for the ball (Randolph et al. 2000, Einav and Randolph 2005). A close linear fit for the ball is Nball-ideal ~ 11.3 + 4. Both sets of results are for a Tresca soil model, and lead to resistance factors for the ball that are 22 to 27 % greater than for the T-bar. This difference reduces using a von Mises strength criterion, for example down to about 15 % for an interface friction ratio of 0.3. Further reduction occurs for anisotropic shear strengths, with a difference of 7 % for a ratio of triaxial extension and compression strengths of 0.5 (Randolph 2000). Experimental data are mixed in relation to any difference between T-bar and ball penetration resistance, with some reported profiles that are indistinguishable (Boylan et al. 2007, Low et al. 2011), whereas profiles in highly sensitive clays show differences of up to 16 %. This difference may be attributed partly to greater reduction in the T-bar resistance due to strain softening, compared with the ball (Einav and Randolph 2005). For soils of moderate sensitivity, the penetration resistances for T-bar and ball are mostly within 5 to 10 %, which is consistent with analytical results that take account of strength anisotropy. In natural soils, as opposed to the idealised perfectly plastic, rate independent material on which plasticity solutions are based, it is essential to allow for the relatively high strain rates in the soil around the penetrometer, and also the gradual softening of the soil as it flows around the cylinder or ball. This has been looked at using a variety of numerical techniques, ranging from a combined upper bound and strain path method (UBSPM; Einav and Randolph 2005), large deformation finite element analysis (LDFE; Zhou and Randolph 2009a), and a steady state finite difference approach (SSFD; Klar and Pinkert 2010). All three approaches adopted a similar logarithmic law of rate dependence, with a relative strength gain of  per tenfold increase in strain rate, and an exponential softening law with 95 % reduction to the fully remoulded shear strength for a cumulative plastic strain of 95. Of the three approaches, the LDFE analysis tends to give the lowest (average) resistance, since it is able to capture the periodic generation and softening of distinct shear bands, accompanied by a corresponding cyclic variation in the penetration resistance. Resistance factors evaluated using LDFE analysis (see Figure 11) may be expressed as (Zhou and Randolph 2009a)

FULL-FLOW PENETROMETERS

Full-flow penetrometers, the cylindrical T-bar and spherical ball (Figure 10), were introduced in the 1990s (Stewart and Randolph 1994, Randolph et al. 1998). The main motivations for their introduction included:  Penetrometer shapes that were amenable to plastic limit analysis, with resistance independent of the pre-yield soil stiffness.  Sufficient ratio of projected area to shaft area to render corrections for pore pressure effects and overburden stress minimal.  Ability to measure remoulded penetration resistance directly, through cycles of penetration and extraction over a limited depth range.  Reduced reliance on site-by-site correlations to obtain resistance factors, and hence shear strength profiles. The last of these has proved something of a disappointment, not helped by an embedded culture with respect to interpretation of cone penetrometer data. Penetrometer is thrust into ground using PROD drill string

Instrumentation, data storage and transmission assembly

Push rod and anti-friction sleeve

  1  4.8 



N Tbar  1  4.8   rem  1   rem e 1.5Tbar / 95 N Tbar  ideal

Spherical ball Pore water pressure filter

N ball

(a) Piezocones and T-bar (b) Ball (Kelleher et al. 2005) Figure 10 Range of penetrometers for in situ testing.

94

rem

 1   rem e

1.5 ball / 95

N

ball  ideal

(26)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013



The quantity, Tbar, in Equation (26) reflects the average plastic shear strain experienced by a typical soil element as it flows around the T-bar. The corresponding quantity for a ball was found to be about 10 % lower (ball ~ 3.3, compared with Tbar ~ 3.7 – Zhou and Randolph 2009a). Of course, the actual degree of softening will vary with the original distance of the soil element from the axis of the advancing penetrometer, since soil elements immediately adjacent to the penetrometer will undergo the greatest softening.

N Tbar ~ 9 1  S t 1

25 No strain softening gradient = 4.8

95 = 50 95 = 25 95 = 15 95 = 10

15

5.1

Parameters  =  rem = 0.2

10

0

0.05

0.1 0.15 Rate parameter 

0.2

Field measurement of consolidation coefficient

The consolidation characteristics of seabed sediments determine the time scale of consolidation following foundation installation, or after cyclic loading that may have caused partial liquefaction. They also determine whether continuous motion, such as a penetrometer test or the axial and lateral movement of a pipeline during thermal buckling, occurs in a drained or undrained manner. It is therefore important to measure the consolidation coefficient, cv, either from laboratory testing or from field dissipation tests following piezocone penetration. Piezocone dissipation tests are commonly interpreted by fitting the excess pore pressure decay to the numerically determined consolidation solution of Teh and Houlsby (1991). This may be approximated (as in Equation (1)) as

Tbar = 3.7

5

(27)

so ranging from a hypothetical 18 for non-softening soil, to a lower limit of 9 for ultra-high sensitivity. For typical sensitivities of offshore sediments in the range of, say, 3 to 10, the resulting resistance factors would lie between 12 and 9.9. Values above or below this range imply respectively higher or lower rate dependency, or sensitivities outside 3 to 10. The form of variation of resistance factor with soil sensitivity is quite similar to that observed experimentally by DeJong et al. (2011) for sensitivities up to about 10, beyond which the experimental resistance factors (based on field vane strength data) continued to fall, with a lower limit of around 6.

NTbar 20



0.25

Figure 11 Values of T-bar resistance factor after allowing for rate effects and strain softening (sensitivity of St = 5, friction ratio,  = 0.2).

The terms, 1 + 4.8, in Equation (26) reflect the average strain rate, which is some 5 orders of magnitude greater than the nominal ‘laboratory’ reference strain rate of 1 %/hr. This term should be viewed with some caution, owing to the limitations of the logarithmic rate law itself, and the inadvisability of trying to extrapolate over such a large range of strain rates. Notwithstanding the above reservation, the analytically derived T-bar and ball factors carry information and should be made use of during the interpretation of field data. Where both ball and T-bar penetrometers are used (and similarly for cone and either T-bar or ball penetrometers), resistance factors should fall within an appropriate relative range, for example with Nball no more than 10 % greater than NTbar unless the soil sensitivity exceeds 10. Low et al. (2010) summarised penetrometer data from a number of offshore (and some onshore) sites around the world, recommending global average resistance factors of 11.9 (with standard deviation of 1.4) for T-bar and ball, relative to an average or laboratory simple shear strength. A similar value of 12 was proposed for NTbar for low sensitivity clays by DeJong et al. (2011), although their ball factor was 10 % higher. These values are plausible, in relation to Figure 11, for example for soils with a rate dependency factor of  ~ 0.1, sensitivity of 3 to 5 and 95 in the range 15 to 25. Some of the parameters that determine the resistance factors can be deduced from the tests themselves; thus cyclic tests enable the sensitivity to be estimated, while tests at different penetration speeds (best performed at the end of a cyclic test when the soil strength has stabilised to the remoulded value) allow the rate parameter to be assessed. The resistance factors from individual sites summarised by Low et al. (2010) suggest that for soils of moderate plasticity the T-bar and ball resistance factors are closer to 11 than 12, while in the ultra-high plasticity soils off the coast of West Africa the average was around 13. This suggests higher strain rate dependency of the West African soils, for example with  closer to 0.15 rather than 0.1. Higher sensitivity implies low interface friction ratio, as well as greater loss of strength during passage of the penetrometer. Numerical analysis for rate dependent ( = 0.1) and softening (95 = 15) material, gave ball resistance factors reducing from 21.5 to 11.6 for sensitivities increasing from 1 to 100 (Zhou and Randolph 2009b). Reducing these by the theoretical ratios for T-bar and ball resistances for Tresca soil leads to a relationship for T-bar resistance factors of:

U

u 1 ~ u ref 1  T / T50 b

(28)

where uref is the reference excess pore pressure that corresponds (ideally) to the initial excess pore pressure at the moment where the piezocone penetration ceases. Time t is normalised as T = cvt/dcone2, and T50 is the normalised time for 50 % excess pore pressure dissipation. (The notation ch is often used, rather than cv, for the consolidation coefficient deduced from piezocone dissipation tests, to emphasise the primary direction of pore fluid flow.) As noted earlier, the exponent, b, is about 0.75, and T50 may be approximated as 0.061 times the square root of the rigidity index, Ir. Determination of cv in this way relies on the penetration phase to have occurred under undrained conditions, for which it is necessary know the consolidation coefficient! Some insight into this circular argument may be obtained by the simple assumption that pore pressure dissipation is a continuous process, some of which may occur during the penetration phase, and the rest of which continues, once the piezocone is halted, during the (subsequent) dissipation phase. This is a slight simplification, but it has proved useful in identifying limits on the reliability of interpreting dissipation tests (DeJong and Randolph 2012). Excess pore pressure data from numerical analysis (e.g. Yi et al. 2012) and experiments (Randolph and Hope 2004, Schneider et al. 2007), where the piezocone was installed at different rates to span drained to undrained conditions, can be fitted by u p0 u ref

~

1 1  V / V50 c

(29)

where up0 is the excess pore pressure during the penetration, which in the field situation would become the initial excess pore pressure for a dissipation test. The normalised velocity, V, is defined as V = vdcone/cv, and V50 is the normalised velocity at which up0 is 50 % of the reference ‘undrained’ excess pore

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pressure, uref. Best fit parameters to the numerical and experimental data are V50 ~ 3, and the exponent c ~ 1. Adopting up0 as the initial excess pore pressure, falsely assuming undrained penetration, will lead to underestimation of the consolidation coefficient, because the time, t50, for (a further) 50 % dissipation will be longer than if the penetration phase had indeed been undrained. Detailed analysis of this is provided by DeJong and Randolph (2012), and the resulting hypothesised relationships between t50 and cv are illustrated in Figure 12, taking Ir = 100 and V50 = 3. An interesting feature of the results is that, for the best fit parameters of b = 0.75 and c = 1, the value of t50 reaches a minimum of about 10 s for cv ~ 400 to 500 mm2/s (around 15,000 m2/yr), before starting to increase again. Obviously this contrasts with the monotonically decreasing relationship for true undrained conditions provided by the Teh and Houlsby (1991) solution.

analyses, to reflect the (primarily) swelling stress path during pore pressure dissipation, giving cv = 19 m2/yr (0.6 mm2/s). The piezocone experimental data match reasonably well the Teh and Houlsby (1991) solution for rigidity index of Ir = 76 (consistent with the model for kaolin adopted in numerical analysis). The experimental piezoball data are compared with a dissipation curve obtained by large deformation finite element (LDFE) analysis using the Modified Cam Clay model and a permeability consistent with the Rowe cell cv value (Mahmoodzadeh et al. 2013). Both theoretical and experimental dissipation curves show a difference in T50 by a factor of 5, compared with the factor of ~2.5 observed by Low et al. (2007) from field tests using a piezoball with pore pressure sensor at the equator. 100 Experimental

Normalised excess pore pressure, u/u0,extrapolated

90

1000 Ir =100, V50 = 3

t50 (s) 100

b = 0.75, c = 1 10 b = 0.75, c = 0.75

b = 1, c = 1

1

100

1000

10000

cv (mm2/s)

cv

78  0.25c  1. 2 v

50 40

Piezoball (mid-face)

Teh & Houlsby (Ir = 76)

30 20

LDFE analysis 0.01 0.1 1 Non-dimensional time factor, T=cvt/d2

10

One of the primary design applications requiring knowledge of the consolidation coefficient is for pipeline design, where the focus is on the upper 0.5 m or so of the seabed. It would be difficult to obtain meaningful data from dissipation testing within that zone, since the proximity to the free surface would affect both the initial stress field following penetration, and potentially the drainage paths and thus the dissipation response. An alternative approach has been proposed recently, which also minimises any time penalty associated with the duration of conventional dissipation testing. The proposed device is a ‘parkable’ piezoprobe, as shown schematically in Figure 14 (Chatterjee et al. 2013). It comprises a solid steel cylinder with hemispherical ends, approximately 250 mm in diameter and 375 mm high. An outrigger may be fitted to provide sufficient force (of 1 to 2 kN) and to limit the embedment to no more than one diameter. The device is designed to be lowered by a winch, from either an ROV or a seabed site investigation system, with measurements gathered in parallel with the main site investigation activities, thus minimising time penalty.

For the same values of V50, b and c, and assuming standard piezocone parameters of dcone = 36 mm and v = 20 mm/s, the relationship in Figure 12 may be written as (DeJong and Randolph 2012): Ir

Piezocone

Figure 13 Dissipation responses from centrifuge model piezocone and piezoball tests compared with numerically derived dissipation curves.

Figure 12 Variation in anticipated t50 with cv, following partially drained penetration.

t 50 ~

Experimental

60

0 0.001

0.1 10

70

10

Teh & Houlsby

1

80

(30)

The corresponding minimum values of t50 range between 7 and 20 s, for rigidity index, Ir, between 50 and 400. From Figure 12, the standard interpretation of a piezocone test becomes questionable once t50 is less than about 50 s. Ball penetrometers are also generally fitted with pore pressure sensors, in commercial practice either at the tip or at the ‘equator’ position (maximum diameter). However, experimental data has shown that, even in normally or lightly overconsolidated clay, the excess pore pressure tends to rise initially at the equator position at the start of a dissipation test, and the overall shape of the dissipation response varies somewhat between tests (DeJong et al. 2008). By contrast, pore pressure measurement at the ‘mid-face’ (a latitude of 45 º south from the equator) gives more consistent data, and with the maximum excess pore pressure occurring at the start of the dissipation test, provided the penetration occurs under undrained conditions (Mahmoodzadeh and Randolph 2013). Typical dissipation responses from centrifuge model tests of piezocone and piezoball penetrometers in normally consolidated kaolin clay are shown in Figure 13. The time axis has been normalised by the diameters (10 mm for the cone, and 15 mm for the ball) and cv values based on data from Rowe cell tests. For the relevant stress level of 110 kPa, the Rowe cell cv is 4 m2/yr. This has been multiplied by the / ratio of 4.7 for the Modified Cam Clay kaolin parameters adopted for the LDFE

Figure 14 Parkable piezoball concept.

In order to provide a theoretical framework to validate the design concept and establish appropriate dissipation curves, LDFE analyses were undertaken. Figure 15 shows contours of

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initial excess pore pressure, normalised by the invert value, for two different boundary conditions (Chatterjee et al. 2013). Natural conditions (left side) with the shear strength increasing linearly with depth were simulated using a nominal 1 kPa surcharge at mudline (the minimum to allow numerical stability during the analysis). Alternatively, in order to simulate approximately uniform strength and stiffness conditions, an artificial 200 kPa surcharge was applied (right side). The resulting consolidation responses are shown in normalised form in Figure 16 for two different embedments (w/D = 0.5 and 1). For comparison, dissipation curves for a pipeline (two-dimensional) and a deeply embedded piezocone (Teh and Houlsby solution) are also shown. The parkable pieozoprobe (PPP) shows more rapid dissipation, for a given diameter, due to its geometry and shallow embedment. For comparison, a curve for a standard diameter piezocone is also shown, scaled according to the relevant diameters. Although the PPP takes longer for the excess pore pressures to dissipate (by a factor of about 7 for t50), the concept for the device is that this occurs in parallel with the primary site investigation activities, so off the critical time line.

export pipelines.) Summaries of recent developments have been provided in the keynote papers of Cathie et al. (2005) and White and Cathie (2010). Here, a brief overview is given of some analytical results that have contributed to design approaches. In deep water, geotechnical design is concerned primarily with issues associated with lateral buckling, which has been the topic of a longstanding joint industry project, the SAFEBUCK JIP (Bruton et al. 2007, 2008). Pipeline buckling is engineered, rather than suppressed, by appropriately spaced buckle initiators, or snake-lay of the pipeline. The axial and lateral resistance offered by the shallow sediments on which the pipelines rest are key inputs to the design. Both of these depend firstly on the embedment of the pipeline into the sediments, and secondly on the velocity and time scale of the movement relative to the soil consolidation characteristics. 6.1

Pipeline embedment occurs during the lay process, while the pipeline is suspended from the lay vessel, in much the same way as a (more permanent) steel catenary riser (SCR) is suspended from a floating production system (Figure 17). Embedment occurs due to the submerged weight of the pipeline, which is augmented by static and dynamic force concentrations for each segment of pipeline as it passes through the touchdown zone. The period within the touchdown zone, and hence the extent of cyclic motions undergone by a given segment of pipe due to wave-induced motions of the lay vessel, will depend on the lay rate; the magnitude of the motions and ratio of dynamic to static force concentration will depend on the sea state conditions as the pipeline is laid. At intermediate depth scales the shear strength profile of deep water sediments may show a mudline intercept of a few kPa (Colliat et al. 2010). However, in the upper 0.5 m that is critical for pipeline design, there is rarely any detectable strength intercept at the mudline. The initial shear strength gradient, , may range from as low as 1 to 1.5 kPa/m, where there is no crustal feature, to ~30 kPa/m, where locally high shear strengths occur, typically at depths of 0.4 to 1 m. Such crustal features are considered to be due to bioturbation (DeJong et al. 2013, Kuo and Bolton 2013).

Excess pore pressure/Invert value 200 kPa surcharge

z/D

1 kPa surcharge

Pipeline embedment

x/D Figure 15 Initial normalised excess pore pressure distributions for cases of strength increasingly linearly with depth (1 kPa surcharge) and quasihomogenous conditions (200 kPa surcharge). 

1.2

Hang‐off point

1

t

0.8

Pipeline: Diameter, D; Bending rigidity, EI Submerged weight, W' T W's  T0 s (arc length) z (constant)

u/ui

Pipeline (w/D = 0.5) 0.6 CPT (using DPPP) CPT (using DCPT)

0.4 PPP, w/D = 0.5 0.2 PPP, w/D = 1 0 0.0001

0.001

0.01

0.1

1

10

zw

100

T = cvt/D2

Seabed (stiff)

(c)

x

Tension, T0

Touchdown point (TDP) Figure 17 Schematic of SCR or pipeline during lay process.

Figure 16 Pore pressure dissipation time history for different geometries and embedment (after Chatterjee et al. 2013).

6

Sea surface

The static penetration resistance for a pipeline of diameter, D, in sediments with strength proportional to depth may be expressed as (Chatterjee et al. 2012a)

PIPELINES AND RISERS

Geotechnical engineering design for pipelines and risers has matured significantly over the last decade, responding to the buckling related design challenges arising from thermal and pressure-induced expansion and contraction of deep water pipelines. (Note, the terminology ‘pipelines’ is used here generically, to include the many different functional names used in the industry, covering flowlines, umbilicals, MEG lines and

w  4 .7   2 D D V

0.17

(31)

where  is the shear strength gradient and V and w are the vertical force per unit length and penetration respectively.

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Within the range of interest (w/D ~ 0.2 to 0.5), this may be approximated by a linear ‘plastic’ spring of stiffness

Normalised seabed stiffness, (kvp/T0)0.5 1

V  k vp ~ 4D w

Maximum contact force, Vmax/W'

(32)

In addition to the ‘geotechnical resistance’ given by the above relationships, allowance must be made for buoyancy effects as the pipeline becomes embedded within the soil (Merifield et al. 2009). This adds a component of resistance that effectively increases the shear strength gradient by a factor that is of the order of 1 + f'/, with f ~ 0.15 to 0.25 depending on the embedment and amount of heave adjacent to the pipeline. Under dynamic lay conditions, lateral motion of the pipe will tend to reduce the amount of soil heave adjacent to the pipe. The linear penetration stiffness allows the complete response of the pipeline (or SCR) to be determined through the touchdown zone, using analytical solutions based on a boundary layer approach (Lenci and Callegari 2005, Palmer 2008, Yuan et al. 2012). A characteristic length, , reflecting the length of the touchdown zone, emerges from the solutions and is given by EI  T0





Figure 18 Profiles of normalised contact force for different values of seabed stiffness.

Westgate et al. (2012) suggested that, as a first approximation, pipeline embedment under typical lay conditions may be estimated based on the maximum static contact force, Vmax, and assuming fully remoulded shear strength for the soil. Combining Equations (32) and (34), but with the shear strength gradient, , replaced by the remoulded strength gradient, rem = /St, then gives



0.25

W'

10000

1

0.3

T0/W' 0.1 OrcaFlex results

10

Analytical solutions (T0/W'  10) Curve fit

Effects of buoyancy may be incorporated by adjusting W' iteratively, or by factoring the remoulded strength gradient by 1 + f'/rem, taking f in the range 0.15 to 0.25. The simple approach of using the remoulded shear strength balances two compensating factors. On the one hand assumption of fully remoulded conditions exaggerates the actual degree of softening under typical lay conditions. This is balanced by using the maximum static force, Vmax, to estimate embedment, rather than the maximum dynamic force, Vdyn, which is typically 25 to 50 % greater than Vmax but can be even larger in more severe sea states (Westgate et al. 2010). A more refined treatment of pipeline embedment was described by Westgate et al. (2013), taking account of:  The estimated number of motion cycles experienced by each section of pipeline as it passes through the touchdown zone.  Combined horizontal and vertical motions.  Gradual softening of the soil resulting from cumulative displacement of the pipeline relative to the soil due to the cyclic motions. The approach builds on the model for cyclic degradation of the resistance of full-flow penetrometers during penetration and extraction cycles (Zhou and Randolph 2009b), but incorporating a brittle ‘structured’ component of soil strength that is lost rapidly (Randolph et al. 2007). The effect of horizontal motion is incorporated by considering theoretical yield envelopes in V-H space, from which an associated flow rule allows estimation of the ratio of vertical to horizontal movements (Cheuk and White 2011). Although built on reasonable theory, the model incorporates empirical adjustment factors, which were calibrated through centrifuge model tests. The model was then applied to three sites where field data were available from post-installation surveys, in addition to video footage during the lay process that allowed estimation of the amplitude of horizontal pipe motions. The observed pipeline embedment was found to lie within the range predicted for ‘light’, ‘moderate’ and ‘severe’ sea states (Figure 20). Direct application of Equation (35), factoring the remoulded shear strength gradient, leads to estimated embedment, w/D, in the range 0.28 to 0.33, which is consistent with the most frequently observed values. However, it is evident from Figure 20, and other field cases reported by Westgate et al. (2013), that the pipeline embedment should be considered as a nondeterministic quantity, varying with lay conditions even if the seabed properties are relatively uniform along the pipeline route. This is consistent with modern probabilistic design approaches for pipelines (White and Cathie 2010). The detailed treatment for estimating pipeline embedment proposed by Westgate et al. (2013) allows probabilistic distributions of pipeline embedment to be derived in a logical manner.

0.25 Vmax ~ 0.6  0.4 2 k vp / T0 (34) W' A typical range for Vmax/W' for deep water pipelines is 1.5 to 3, as indicated in Figure 18.



1000

100

where EI is the bending rigidity of the pipe and T0 the horizontal component of tension in the catenary (Figure 17). The effect of the seabed stiffness, kvp, on the profile of contact force, V, through the touchdown zone is shown in Figure 18. The maximum static contact force, Vmax, normalised by the submerged weight of the pipe, W', is a function of the seabed stiffness, and also of the characteristic length, , as shown in Figure 19. The variation of Vmax may be approximated as (Randolph and White 2008a)

w 1  2 rem D / T0 ~ D 7 rem D2

100

Figure 19 Maximum static contact force in touchdown zone (Randolph and White 2008a).

(33)



10

1

(35)

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on the trailing edge of a pipeline is extremely brittle, so that the relevant failure envelope reverts quickly to that for no tension. The theoretical failure envelopes referred to above are based on ideal, rate independent non-softening soil, and ignore any local heave (relative to the flat seabed) adjacent to the pipeline. A more realistic study, based on LDFE analysis, that takes account of such features was presented by Chatterjee et al. (2012b). For the particular set of soil parameters investigated, they derived failure envelopes that were approximately parabolic, expressed as



H max Vmax

 V   Vmax

  

1

 V 1  V max 

  

2

   2 1 2 where  1



H  0 Vmax (36)



1 1  22

Figure 20 Comparison of predicted and observed pipeline embedment (from Westgate et al. 2013, Site C).

6.2

The quantities 1, 2 and Hmax/Vmax were found to vary with embedment according to

Lateral resistance

1  0.59  0.89 w / D

The lateral resistance of partially embedded pipelines may be assessed, as for shallow foundations, from failure surfaces in vertical (V) – horizontal (H) load space. The form and size of failure envelope depend on factors such as the embedment, the pipe-soil interface condition (friction ratio, , ranging between 0 for smooth to 1 for rough; and whether tensile stresses are permitted), the shear strength profile (ranging from uniform to varying proportionally with depth) and the relative magnitudes of effective stress and shear strength. Theoretical failure envelopes considering some or all of these variables have been presented by Randolph and White (2008b: analytical upper bound solutions), Merifield et al. (2008: finite element analyses) and Martin and White (2012: closely bracketed finite element based lower and upper bound plasticity solutions). An example from the most recent of these is shown in Figure 21, for a fully rough pipeline embedded in soil with strength proportional to depth, for two different ratios of effective stress to shear strength gradient ('/). '/ = 0

'/ = 3

H/D2

H/D2

 2 0.55  0.87 w / D

(37)

H max / Vmax 0.17  0.31 w / D

 The failure envelopes allow estimation of the breakout lateral resistance for any given vertical load ratio, V/Vmax, and loading path. As a pipeline is displaced laterally it tends to rise towards the seabed, or plunge deeper, depending on the initial embedment and vertical load ratio. After sufficient movement it will reach a steady residual horizontal resistance, Hres. Pipeline trajectories during breakout, and a methodology for assessing the residual resistance ratio, Hres/V, were also presented by Chatterjee et al. (2012b). 6.3

Axial resistance

The axial resistance of pipelines is an intriguing problem that, at face value, would seem to be essentially trivial (a sliding failure with known vertical load), but in practice turns out to be more complex. There are three main aspects that need to be considered (Hill et al. 2012):  The pipe-soil interface friction, which is affected by the relative roughness of the pipeline coating, and also the magnitude of the normal effective stress. At the very low effective stresses (generally less than 5 kPa) applied by deep water pipelines, the effective stress failure envelope shows significant curvature.  The cylindrical geometry of the pipeline, which for any given embedment leads to integrated normal effective stresses around the pipe-soil interface that exceed the pipeline weight by a so-called ‘wedging factor’.  Excess pore pressure development at the pipe-soil interface, which leads to a strong dependence of the axial resistance on the velocity and cumulative axial displacement. The first of these requires appropriate experimental data, typically obtained using direct shear devices that have been adapted for very low normal stresses. Analytical solutions can provide a theoretical basis for the other aspects, and these are discussed here. The basis for estimating the wedging factor, , due to the cylindrical pipe surface is illustrated in Figure 22. Drawing on the classical solution for the stresses due to a line load acting on the surface of a homogeneous elastic half-space, a cos  variation of the normal effective stress may be assumed (with the magnitude of the induced ‘radial’ stresses decaying inversely with radius from the pipe centre). Integrating the

V/D2

Figure 21 Examples of failure envelopes for rough pipelines in soil with strength proportional to depth (Martin and White 2012).

There is a significant difference in lateral and uplift resistance depending on the assumption of full tension or no tension at the pipeline surface. The slight uplift resistance for the case of no tension is primarily due to soil above the pipeline for embedment ratios exceeding 0.5. In practice, model test data indicate that, during lateral displacement, the tensile resistance

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normal effective stresses around the perimeter, P, of the interface, with S adjusted to balance the submerged pipeline weight, W', allows the average normal effective stress, q, to be expressed as 2 sin  m W with    1.27 P  m  sin  m cos  m

 F  F  F   ln( 2 ) T / T50 m  F    e      W '  W '  d  W '  d  W '  u 

(38)

and P D m  D/2

The value of m is related to the normalised embedment by cos m = 1 - 2w/D.

Pipe diameter D

0.7

m

W'

0.6



w

P

n' ~ Scos

Figure 22 Schematic of embedded pipe.

Within a conventional critical state framework, the effect of the time scale for axial movement may be evaluated by considering the tendency for the adjacent soil to compact, or dilate, and the resulting maximum (in an absolute sense) excess pore pressure that may develop during rapid shearing. This is illustrated in Figure 23 for the case of contractive soil. The potential maximum reduction in void ratio, -emax, during drained shearing is equivalent to a state parameter (Been and Jefferies 1985), although more usefully expressed in terms of volumetric strain, v,max. The corresponding maximum excess pore pressure during undrained shearing is then obtained from   u max ln1  q 

 v ,max 1  e0  e max      v ,max   *   

umax e 

u emax critical state line q

Pipe w/D = 0.4 Backbone curve fit

0.5

0.0039

0.4 0.3

0.039 390 0.39

0.2

39

3.9

Values of vD/cv as indicated

0.1 0 0.00001 0.0001

0.001

0.01

0.1

1

10

100

cvt/D2

Figure 24 Example axial response of pipeline as a function of nondimensional time and velocity.

Unfortunately, data from extensive model tests on pipe segments are not consistent with the theoretical framework of critical state soil mechanics and consolidation described above. The data show that excess pore pressures may be regenerated during fast axial motion that follows slow, drained, axial displacement, and indeed the axial friction has generally been considered as a function of the shearing velocity rather than the elapsed time during a given motion (White et al. 2011). A model that broadly reproduces the trends observed in the model tests was suggested by Randolph et al. (2012). The model supposes that pore pressure is continuously generated during shearing, in response to volumetric collapse (generically referred to as ‘damage’) within the soil adjacent to the pipe. The rate of (potential) volumetric strain was assumed proportional to the shear strain rate (denoted by normalised velocity, v/D) and to the current normal effective stress, so that no further damage would occur if the effective stress were to fall to zero. Taking the rate of volumetric collapse (or damage) as v/D, the rate of excess pore pressure generation becomes

(39)

e e0

(40)

where the subscripts d and u denote drained and undrained limits, m ~ 0.5 and T50, representing the non-dimensional time where the friction ratio is midway between drained and undrained limits, is about 0.05. 

Mobilised axial friction, F/W'

q 

dissipate and the friction ratio increases to the drained value. The form of the backbone curve that quantifies the degree of consolidation as a function of T = cvt/D2 may be approximated as (Randolph et al. 2012): 

ln 'n

Figure 23 Critical state framework for stress paths during shearing.

The proportion of umax that develops at the pipeline-soil interface depends on the velocity and time scale (or cumulative displacement) of the axial motion. For slow movement, excess pore pressure can dissipate as fast as it is generated, and the response is fully drained, while at the opposite extreme high excess pore pressures are generated initially, although should dissipate with continued displacement. An example response is shown in Figure 24 from FE analysis of a pipeline resting on normally consolidated Modified Cam Clay, with a plane strain friction angle of 27 º (Randolph et al. 2012). The theoretical wedging factor for w/D = 0.4 is 1.25, so that the drained axial friction factor is F/W' ~ 1.25tan(27) = 0.64. For fast shearing the initial excess pore pressure ratio, umax/q, is about 0.45, so that the undrained friction ratio is (1 – 0.45)×0.64 = 0.35. With increasing elapsed time, or displacement (noting that cvt/D2 is equivalent to (/D)/(vD/cv)), the excess pore pressures

100

u   v du / dt 1     * D  q  q

(41)

At high rates of shearing, the effect of damage is partially compensated by slight enhancement of the effective friction ratio due to increased shear strain rates. This may be modelled using standard models for rate dependency of shear strength, for example a form of Herschel-Bulkley relationship, so that the failure shear stress ratio becomes   v/D     f u  u       y 1   1     HB 1   q q  q     v ref / D   

(42)

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where y is the minimum (yield) friction coefficient at very slow shearing rates and  and  are the rate parameters; these combine to give the rate-enhanced friction coefficient, HB. Using the backbone consolidation curve shown in Figure 24 (Equation (40)) as the basis for pore pressure dissipation, the excess pore pressure may be obtained by a convolution integral of the form t  v( t ' )  ln 2T  T'  / T50 m u  dt'   * D q  u(t' )e t' 0

(43)

where v and u are both time varying functions and T' = cvt'/D2. An example response is shown in Figure 25. Of particular note is that after an initial transient stage, the normalised friction, /q, converges to a steady value that is a function of velocity. At steady state, pore pressure generation due to damage balances pore pressure dissipation due to consolidation. The steady state friction was approximated as  steady state   1    HB  1  q  1  0.24 /  /  *T50 vD / c v  

k a  mD

Figure 25 Example axial response of pipeline incorporating damage and strain rate (Randolph et al. 2012).

(46)

For a partially embedded pipeline, this may be reduced by a factor sin m, where m is defined in Figure 22. By comparison, the vertical stiffness for a (surface) foundation of width Dsin m on similar soil would be given by kv = 2mDsin m (Gibson 1974). Hence the axial stiffness is about half the vertical stiffness (a little lower, allowing for the embedded nature of the pipeline, Guha 2013). The axial load transfer stiffness may be combined with the expression for the stiffness of a long pile (Equation (8)) in order to yield the overall pipeline stiffness for axial motion:

K pipe, axial  6.4

(44)

Although this model of velocity and time-dependent axial friction contains some speculative elements, such as the proposed link between pore pressure generation and normalised velocity, it provides a theoretical framework for design, and for the planning of future model tests in the laboratory or field. It also helps to resolve the apparent discrepancy between conventional consolidation theory and experimental data.

EApipe k a

EApipe mD



(47)

Impact forces from debris flows

Geohazard assessment, particularly from submarine landslides, is a major aspect of developments in deep water, i.e. beyond the continental shelf, where relic landslides are frequently observed. While it is generally possible to site well manifolds and anchoring systems away from the flow paths of potential landslides, pipelines (particularly export pipelines) by their nature must frequently be exposed to some risk. It is therefore necessary to consider the magnitude of impact forces from debris flows, and also the resulting response of a pipeline in order to gauge whether it would survive impact. The problem to be considered is shown schematically in Figure 26. The debris flow may be idealised as extending over a finite width, B, within which it imparts a normal force (per unit length), Fn, and an axial force, Fa. Outside the impact zone, passive lateral and axial resistance is provided between the pipeline and the soil. Generic analytical solutions have been developed for the pipe response for given non-dimensional ratios of active loading to passive resistance, allowing estimates of the maximum stresses induced in the pipeline and maximum deflection under the action of the debris flow (Randolph et al. 2010). However, methods to estimate the loading itself have tended to lack a sound fundamental basis, being couched in terms of drag factors for normal and parallel components of flow. These lead to resistances that are functions of density and velocity of flow, rather than parameters linked to shear strength or even viscosity. Flow direction

Axial stiffness

Debris flow

In addition to evaluating the limiting pipe-soil friction ratio, the pre-failure axial stiffness of the pipeline is important as a boundary condition for analysis of pipeline walking or the feedin to lateral buckles or debris flow impact. At an element level, the axial stiffness (ratio of load transfer per unit length to axial displacement) may be estimated by assuming a simple distribution of shear stress around the perimeter of the pile, similar to that for normal effective stress (Figure 22). Consider a pipeline that is embedded to w/D = 0.5, and where the shear stress resisting axial movement varies as cos  around the embedded section of the pipe. The shear stress will also decrease inversely with radius from the pipe axis, in order to satisfy equilibrium. Now assume a shear modulus for the soil that varies proportionally with depth, z, according to G = mz. At any radial position, the shear strain will therefore be

  D cos   inv D    inv  G 2r 2 m cos  2r 2 m

where inv is the shear stress at the pipe invert. Integrating this with respect to r leads to the displacement at the pipe. The resulting axial load transfer stiffness is then given by

(45)

101

Fn

Pipeline

 Fa Passive region resisting movement

Active region loaded by slide

Passive region resisting movement

Figure 26 Schematic of debris flow impacting pipeline (Randolph and White 2012).

For flow normal to the pipeline ( = 90 º in Figure 26) a hybrid approach, combining ‘geotechnical’ and ‘fluid drag’ components of resistance, was proposed by Randolph and White (2012). The normal force per unit length of pipe, Fn, is expressed as

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(48)

where Np is a bearing factor, su,op is the operative shear strength at a shear strain rate that reflects the (normal component of) flow velocity, vn, and Cd is a drag coefficient. The relationship was calibrated against numerical analysis data (Zakeri 2009), and yielded drag coefficients in the range 0.6 to1.2 for flow angles between 30 and 90 º. The principle behind Equation (48) is that the bearing factor, Np, in common with other bearing factors in geotechnics, captures the geometry of the failure mechanism, and should be independent of velocity or soil strength, essentially as specified in Equation (13) but with adjustment for the relative depth of the debris flow compared with the pipeline diameter. The effect of velocity, or shear strain rate, is incorporated into the operative shear strength, using conventional relationships such as the Herschel-Bulkley expression in Equation (42), or a simple power law relationship:

  s u,op  s u,ref    ref



 v /D    ~ s u,ref  n   ref  

between debris flow and pipeline. Initially, as the debris flow strikes the pipeline, it will carry the pipe with it. Resisting bending moments and axial tension in the pipeline will develop quite gradually as the pipeline is deformed. These will slow the pipeline, relative to the debris flow, until a dynamic equilibrium is established (Boylan and White 2013). A single set of results from Randolph et al. (2010) is shown in Figure 28, for a case where the passive horizontal resistance of the pipeline outside the slide zone is half the active force, Fn, and the passive axial resistance is 25 % of Fn. The total active loading, Fn times the slide zone width B, is normalised by the pipeline cross-sectional rigidity, EA. The strains in the pipe become dominated by axial tension as the width of the debris flow increases; it is evident that relatively low levels of active loading can cause significant strains, and potentially failure of the pipeline. 1.6



(49)

The relative magnitudes of the two components in Equation (48) are such that the fluid drag term only becomes significant once the Johnson number (also referred to as the nonNewtonian Reynolds number), vn2/su,op exceeds about 5. The accuracy of this approach has recently been demonstrated through experimental work (Sahdi et al. 2013), where a drag factor of around 1.1 to 1.4 was suggested. Numerical analyses using the material point method (Ma, private communication) has confirmed a drag factor close to unity. For flow parallel to the pipeline, analytical relationships have been derived for material that follows a power law function, as in Equation (49) (Einav and Randolph, 2006). The axial force per unit length, Fa, is given by

0º

1.4 Axial coefficients, f a

Fn 1  N p s u,op  Cd  v 2n   D 2  

1.2 Failure envelope 1

30 º

Relative angle between debris flow and pipelines 45 º

0.8 60 º 0.6 0.4 0.2 90 º

0 0

5

10

15

Normal coefficients, Np Figure 27 Failure envelope for varying flow angle relative to pipe axis (Randolph and White 2012). 0.01

  1   Fa f a s u ,op D where f a   2   1      

(50)

The value of fa lies in the range 1.2 to 1.4 for typical values of  between 0.05 and 0.15. For the general case of debris flow impacting a pipeline at an angle , a failure envelope may be developed to quantify the interaction between parallel and normal components of force. Based on the numerical data from Zakeri (2009), a failure envelope of the form  fa   f a ,0 

3

  Np     N p,90  

1

  1  

with N p  N p,90 sin 0.7

(51)

Maximum pipeline strain, /E

Combined



102

0.0005 0.001

0.0002 0.0001 0.00005 0.00002

Tension 0.0001

Bending 0.00001 10

100

1000

10000

Normalized debris flow width, B/D

Figure 28 Effect of slide loading and width on maximum pipeline strain (Randolph et al. 2010).

7

was found to give a reasonable fit (Randolph and White 2012). An example failure envelope, taking fa,0 = 1.4 and Np,90 = 11.9 as appropriate for a rough pipe, is shown in Figure 27, together with spot points for flow angles of 0, 30, 45, 60 and 90 º. Assessment of pipeline response to debris flow impact requires initial estimation of debris flow velocity, height (which affects Np), relative angle and shear strength at the point of impact. These are non-trivial quantities to estimate, but may be gleaned from numerical modelling of landslide runout. The resulting impact forces and pipeline response may then be evaluated using the relationships summarised here. An important consideration is that the normal velocity, vn, used to determine the strain rate (hence operative shear strength) and the drag force should be the relative velocity

FnB/EA = 0.001

CONCLUSIONS

Analysis underpins and enriches design approaches that we use in day to day practice. Where empirical correlations are still relied upon, we should strive continuously to understand the underlying processes and gradually capture them quantitatively through analysis or synthesis of well-considered numerical studies. The paper has dipped into a number of different application areas in offshore geotechnical design, with the aim throughout being to present simplified outcomes, based on analysis, that can be applied directly in design. It should be emphasised, however, that simplifications and idealisations in analytical solutions are such that final validation and fine-tuning of a design will often require further input from physical or numerical modelling of the specific application. Even there though, analytical solutions should guide the planning of the more sophisticated investigations.

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Perhaps most importantly, analytical solutions are the clearest language through which engineering systems educate us in respect of the controlling behaviour in offshore geotechnical design. Simple relationships indicate which parameters we should pay close attention to and which parameters have less influence. In the early stages of a project, analytical solutions can highlight the parameters that are most important when targeting site investigations, and which aspects of our design offer the most scope for optimising performance. 8

ACKNOWLEDGEMENTS

The work reported here is underpinned by the activities of the Centre for Offshore Foundation Systems (COFS), currently supported as a node of the Australian Research Council Centre of Excellence for Geotechnical Science and Engineering, and in partnership with The Lloyd’s Register Educational Trust. Support through the Australian Research Council’s Discovery Program is also acknowledged. However, the most important acknowledgement is for the many colleagues in COFS, Advanced Geomechanics, and elsewhere who have contributed to specific results and any useful ideas presented here. 9

REFERENCES

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Gourvenec, S. 2007a. Shape effects on the capacity of rectangular footings under general loading. Géotechnique 57(8), 637-646. Gourvenec, S. 2007b. Failure envelopes for offshore shallow foundation under general loading. Géotechnique 57(9), 715-727. Gourvenec, S. 2008. Undrained bearing capacity of embedded footings under general loading. Géotechnique 58(3), 177–185. Gourvenec, S. and Barnett, S. 2011. Undrained failure envelope for skirted foundations under general loading. Géotechnique 61(3), 263–270. Gourvenec, S. and Randolph, M.F. 2003. Effect of strength nonhomogeneity on the shape and failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique 53(6), 575-586. Guha, I. 2013. Structural analysis of submarined pipelines under submarine slide and thermal loading. Forthcoming PhD thesis, University of Western Australia. Guo, W.D. and Randolph, M.F. 1997. Vertically loaded piles in nonhomogeneous media. Int. J. Num. and Anal. Methods in Geomechanics 21(8), 507-532. Hill, H.J., White, D.J., Bruton, D.A.S., Langford, T., Meyer, V., Jewell, R.A. and Ballard, J.-C. 2012. A new framework for axial pipe-soil resistance illustrated by a range of marine clay datasets. Proc. 7th Int. Conf. Offshore Site Investigation and Geotechnics, Soc. for Underwater Technology, London, 367-377. Hodder, M.S., White, D.J. and Cassidy, M.J. 2013. An effective stress framework for the variation in penetration resistance due to episodes of remoulding and reconsolidation. Géotechnique 63(1), 30–43. Houlsby, G.T. and Puzrin, A.M. 1999. The bearing capacity of a strip footing on clay under combined loading. Proc. R. Soc. London A 455, 893–916. ISO 2003. ISO 19901-4: Petroleum and natural gas industries — Specific requirements for offshore structures — Part 4: Geotechnical and foundation design considerations, 1st Edition. International Standards Organisation, Geneva. ISO 2007. ISO 19902: Petroleum and natural gas industries — Fixed steel offshore structures, 1st Edition. International Standards Organisation, Geneva. Jardine, R.J., Chow, F.C., Overy, R.F. and Standing, J.R. 2005. ICP design methods for driven piles in sands and clays. Telford, London. Jeanjean, P. 2006. Set-up characteristics of suction anchors for soft Gulf of Mexico clays: experience from field installation and retrieval. Proc. Offshore Technology Conf., Houston, Paper OTC 18005. Jeanjean, P. 2009. Re-assessment of p-y curves for soft clays from centrifuge testing and finite element modeling. Proc. Offshore Technology Conf., Houston, Paper OTC 20158. Jeanjean, P. 2012. State of practice: Offshore geotechnics throughout the life of an oil and gas field. Proc. GeoCongress 2012, State of the Art and Practice in Geotechnical Engineering, Oakland, Ca, ASCE Geotechnical Special Publication No. 226, 643-677. Jeanjean, P., Watson, P.G., Kolk, H. and Lacasse, S. 2010. RP 2GEO: The new API recommended practice for geotechnical engineering. Proc. Offshore Technology Conf., Houston, Paper OTC 20631. Kelleher, P.J. and Randolph, M.F. 2005. Seabed geotechnical characterisation with the portable remotely operated drill. Proc. Int. Symp. on Frontiers in Offshore Geotechnics, ISFOG, Perth, 365-371. Klar, A. and Pinkert, M.F. 2010. Steady-state solution for cylindrical penetrometers. Int. J. Num. and Anal. Methods in Geomechanics 34, 645-659. Klar, A. and Randolph, M.F. 2008. Upper bound and load displacement solutions for laterally loaded piles in clay based on energy minimisation. Géotechnique 58(10), 815-820. Kraft, L.M., Ray, R.P. and Kagawa, T. 1981. Theoretical t-z curves. J. Geot. Eng. Div., ASCE, 107(11), 1543-1561. Kuo, M.Y-H. and Bolton, M.D. 2013. The nature and origin of deep ocean clay crust from the Gulf of Guinea. Géotechnique in press. Lehane, B.M., Schneider, J.A. and Xu, X.2005. The UWA-05 method for prediction of axial capacity of driven piles in sand. Proc. Int. Symp. on Frontiers in Offshore Geomechanics ISFOG, Perth, 683– 689. Lenci, S. and Callegari, M. 2005. Simple analytical models for the J-lay problem, Acta Mechanica 178, 23-39. Low, H.E., Landon, M.M., Randolph, M.F. and DeGroot, D.J. 2011. Geotechnical characterisation and engineering properties of Burswood clay. Géotechnique 61(7), 575-591. Low, H.E., Lunne, T., Andersen, K.H., Sjursen, M.A., Li, X. and Randolph, M.F. 2010. Estimation of intact and remoulded

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undrained shear strength from penetration tests in soft clays. Géotechnique 60(11), 843-859. Low, H..E., Randolph, M.F. and Kelleher, P. 2007. Estimation of in-situ coefficient of consolidation from dissipation tests with different penetrometers. Proc. 6th Int. Conf. Offshore Site Investigation and Geotechnics, Soc. for Underwater Technology, London, 547-556. Mahmoodzadeh, H. and Randolph, M.F. 2013. The effect of partial consolidation on the subsequent dissipation test. Under review. Mahmoodzadeh, H., Wang, D. and Randolph, M.F. 2013. Interpretation of piezoball dissipation test in kaolin clay. Under review. Martin, C.M. 2001. Vertical bearing capacity of skirted circular foundations on Tresca soil. Proc. 15th Int. Conf. on Soil Mechanics and Geotechnical Engineering, Istanbul, 1, 743-746. Martin, C.M. and Randolph, M.F. 2006. Upper bound analysis of lateral pile capacity in cohesive soil. Géotechnique 56(2), 141-145. Martin, C.M. and White, D.J. 2012. Limit analysis of the undrained bearing capacity of offshore pipelines. Géotechnique 62(9), 847863. Merifield, R.S., White, D.J. and Randolph, M.F. 2008. The ultimate undrained resistance of partially-embedded pipelines. Géotechnique 58(6), 461-470. Merifield, R.S., White, D.J. and Randolph, M.F. 2009. The effect of surface heave on the response of partially-embedded pipelines on clay. J. Geotech. Geoenviron. Eng., ASCE, 135(6), 819-829 Murff, J.D. 1975. Response of axially loaded piles. J. Geot. Eng. Div., ASCE 101(GT3), 356-360. Murff, J.D. 1980. Pile capacity in a softening soil. Int. J. Numerical and Analytical Methods in Geomechanics 4, 185–189. Murff, J.D. 1994. Limit analysis of multi-footing foundation systems. Proc. 8th Int. Conf. on Computer Methods and Advances in Geomechanics, Morgantown, 1, 223-244. Murff, J.D. 2012. Inaugural McClelland lecture: Estimating the capacity of offshore foundations. Proc. 7th Int. Conf. Offshore Site Investigation and Geotechnics, Soc. for Underwater Technology, London, 9-44. Murff, J.D., Aubeny, C.P. and Yang, M. 2010. The effect of torsion on the sliding resistance of rectangular foundations. Proc, 2nd Int. Symp. Frontiers in Offshore Geotechnics, ISFOG 2010, Perth, 439443. Murff, J.D. and Hamilton, J.M. 1993. P-ultimate for undrained analysis of laterally loaded piles. J. Geot. Eng. Div., ASCE, 119(1), 91-107. Mylonakis, G. and Gazetas, G. 1998. Settlement and additional internal forces of grouped piles in layered soil. Géotechnique 48(1), 55–72. Neubecker, S.R. and Randolph, M.F. 1995. Profile and frictional capacity of embedded anchor chain. J. Geot. Eng. Div., ASCE, 121(11), 787-803. Neubecker S.R. and Randolph M.F. 1996. The performance of drag anchors and chain systems in cohesive soil. Marine Georesources and Geotechnology 14, 77-96. Novello, E.A. 1999. From static to cyclic p-y data in calcareous sediments. Proc. 2nd Int. Conf. on Engineering for Calcareous Sediments, Bahrein, 1, 17–24. O'Neill, M.P., Bransby, M.F. and Randolph, M.F. 2003. Drag anchor fluke-soil interaction in clay. Canadian Geotechnical J. 40(1), 7894. Palmer A.C. 2008. Touchdown indentation of the seabed. Applied Ocean Research 30, 235-238. Poulos H.G. 1988 Cyclic stability diagram for axially loaded piles. J. Geotech. Geoenviron. Eng., ASCE, 114 (8), 877-895. Puech, A., Benzaria, O., Thorel, L., Garnier, J., Foray,P., Silva, M. and Jardine, R.J. 2013. Diagrammes de stabilité cyclique de pieux dans les sables. Proc. 18th Int. Conf. on Soil Mechanics and Geotechnical Engineering, Paris. Randolph, M.F. 1983. Design considerations for offshore piles. Proc. Conf. on Geot. Practice in Offshore Engng, ASCE, Austin, 422-439. Randolph M.F. 2000. Effect of strength anisotropy on capacity of foundations. Proc. John Booker Memorial Symp., Sydney, 313-328. Randolph, M.F. 2003. 43rd Rankine Lecture: Science and empiricism in pile foundation design. Géotechnique 53(10), 847-875. Randolph, M.F., Hefer, P.A., Geise, J.M. and Watson, P.G. 1998. Improved seabed strength profiling using T-bar penetrometer. Proc Int. Conf. Offshore Site Investigation and Foundation Behaviour, Society for Underwater Technology, London, 221-235. Randolph, M.F. and Hope, S. 2004. Effect of cone velocity on cone resistance and excess pore pressures. Proc. Int. Symp. On Eng. Practice and Performance of Soft Deposits, Osaka, 147-152.

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Randolph, M.F. and Houlsby, G.T. 1984. The limiting pressure on a circular pile loaded laterally in cohesive soil. Géotechnique 34(4), 613-623 Randolph, M.F., Low, H.E. and Zhou, H. 2007. In situ testing for design of pipeline and anchoring systems, Proc. 6th Int. Conf. on Offshore Site Investigation and Geotechnics, Soc. for Underwater Technology, London, 251-262. Randolph, M.F., Martin, C.M. and Hu, Y. 2000. Limiting resistance of a spherical penetrometer in cohesive material. Géotechnique 50(5) 573-582. Randolph, M.F. and Puzrin, A.M. 2003. Upper bound limit analysis of circular foundations on clay under general loading. Géotechnique 53(9), 785-796. Randolph, M.F., Seo, D. and White, D.J. 2010. Parametric solutions for slide impact on pipelines. J. Geotech. Geoenviron. Eng., ASCE, 136(7), 940-949. Randolph, M.F. and White, D.J. 2008a. Pipeline embedment in deep water: processes and quantitative assessment. Proc. Offshore Technology Conf., Houston, Paper OTC 19128. Randolph, M.F. and White, D.J. 2008b. Upper bound yield envelopes for pipelines at shallow embedment in clay. Géotechnique 58(4), 297-301. Randolph, M.F. and White, D.J. 2012. Interaction forces between pipelines and submarine slides – a geotechnical viewpoint. Ocean Engineering 48, 32-37. Randolph, M.F., White, D.J. and Yan, Y. 2012. Modelling the axial soil resistance on deep water pipelines. Géotechnique 62(9), 837-846. Randolph, M.F. and Wroth, C.P. 1978. Analysis of deformation of vertically loaded piles. J. Geot. Eng. Div., ASCE, 104(GT12), 1465-1488. Randolph, M.F. and Wroth, C.P. 1979. An analytical solution for the consolidation around a driven pile. Int. J. Num. and Anal. Methods in Geomechanics 3(3), 217-229. Sahdi, F., Gaudin, C., White, D.J., Boylan, N. and Randolph, M.F. 2013. Centrifuge modelling of active slide-pipeline loading in soft clay. Géotechnique (under review). Salgado R, Lyamin A.V., Sloan S.W. and Yu H.S. 2004. Two and threedimensional bearing capacity of foundations in clay. Géotechnique 54(5), 297-306. Schneider, J. A., Lehane, B. M., and Schnaid, F. 2007. Velocity effects on piezocone tests in normally and overconsolidated clays. Int. J. Physical Modelling in Geotechnics 7(2), 23–34. Schneider, J.A., Xu, X. and Lehane, B.M. 2008. Database assessment of CPT based design methods for axial capacity of driven piles in siliceous sands. J. Geotech. Geoenviron. Eng., ASCE, 134(9), 1227-1244. Stewart, D.P. and Randolph, M.F. 1994. T-Bar penetration testing in soft clay. J. Geot. Eng. Div., ASCE 120(12), 2230-2235. Suryasentana, S.K. and Lehane, B.M. 2013. Numerical derivation of CPT-based p-y curves for piles in sand. Géotechnique, under review. Taiebat, H.A. and Carter, J.P. 2000. Numerical studies of the bearing capacity of shallow foundations on cohesive soil subjected to combined loading. Géotechnique 50(4), 409-418. Taiebat, H.A. and Carter, J.P. 2002. Bearing capacity of strip and circular foundations on undrained clay subjected to eccentric loads. Géotechnique 52(1), 61-64. Taiebat H.A., and Carter, J.P. 2010. A failure surface for circular footings on cohesive soils. Géotechnique 60(4), 265–273. Teh, C.I. and Houlsby, G.T. 1991. An analytical study of the cone penetration test in clay. Géotechnique 41(1), 17–34. Tian, Y., Cassidy, M.J., Gaudin, C. and Randolph, M.F. 2013. Considerations on the design of keying flap of plate anchors. J. Geotech. Geoenviron. Eng., ASCE, in press. Wei, Q., Cassidy, M.J., Tian, Y. and Gaudin, C. 2013. Incorporating shank resistance into prediction of the keying behaviour of suction embedded plate anchors. Under review.

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Wesselink, B.D., Murff, J.D., Randolph, M.F., Nunez, I.L. and Hyden, A.M. 1988. Analysis of centrifuge model test data from laterally loaded piles in calcareous sand. Proc. Int. Conf. on Engineering for Calcareous Sediments, Perth, 1, 261-270. Westgate, Z.J., Randolph, M.F., White D.J. and Li, S. 2010. The influence of sea state on as-laid pipeline embedment: a case study, Applied Ocean Research 32(3), 321-331. Westgate, Z., White, D.J. and Randolph, M.F. 2012. Field observations of as-laid pipeline embedment in carbonate sediments. Géotechnique 62(9), 787-798. Westgate, Z., White, D.J. and Randolph, M.F. 2013. Modelling the embedment process during offshore pipe laying on fine-grained soils. Canadian Geotechnical Journal, in press. White D.J., Bolton M.D., Ganesan S.A., Bruton D.A.S., Ballard J.-C. and Langford T. (2011). SAFEBUCK JIP: Observations from model testing of axial pipe-soil interaction on soft natural clays. Proc. Offshore Technology Conf., Houston, Paper OTC 21249. White, D.J. and Cathie, D.N. 2010. Geotechnics for subsea pipelines. Proc. 2nd Int. Symp. on Frontiers in Offshore Geotechnics, ISFOG 2010, Perth, 87-123. White, D.J. and Lehane, B.M. 2004. Friction fatigue on displacement piles in sand. Géotechnique 54(10), 645–658. White, D.J., Schneider, J.A. and Lehane, B.M. 2005. The influence of effective area ratio on shaft friction of displacement piles in sand. Proc. Int. Symp. on Frontiers in Offshore Geomechanics, ISFOG, Perth, 741–747. Yang, M., Aubeny, C.P. and Murff, J.D. 2010. Undrained capacity of plate anchors under general loading. J. Geotech. Geoenviron. Eng., ASCE 136(10), 1383-1393. Yang, M., Aubeny, C.P. and Murff, J.D. 2012. Behaviour of suction embedded plate anchors during the keying process. J. Geotech. Geoenviron. Eng., ASCE 138(2), 174–183. Yi, J.T., Goh, S.H., Lee, F.H. and Randolph, M.F. 2012. A numerical study of cone penetration rate effects, Géotechnique 62(8),707-719. Yuan F., Wang L., Guo, Z. and Xie Y.G. 2012. Analytical analysis of pipeline-soil interaction during J-lay on a plastic seabed with bearing resistance proportional to depth. Applied Ocean Research 36, 60-68. Yun, G. and Bransby, M.F. 2007. The undrained vertical bearing capacity of skirted foundations. Soils and Foundations 47(3), 493505. Yun, G.J., Maconochie, A., Oliphant, J. and Bransby, M.F. 2009. Undrained capacity of surface footings subjected to combined V-HT loading. Proc. Int. Offshore and Polar Engineering Conference, Osaka, Paper 2009-TPC-614. Zakeri, A. 2009. Submarine debris flow impact on suspended (freespan) pipelines: normal and longitudinal drag forces. Ocean Engineering 36(6-7), 489-499. Zakeri, A., Chi, K. and Hawlader, B. 2011. Centrifuge modeling of glide block and out-runner block impact on submarine pipelines. Proc. Offshore Technology Conf., Houston, Paper OTC 21256. Zhang, C., White, D.J., Randolph, M.F. 2011. Centrifuge modelling of the cyclic lateral response of a rigid pile in soft clay. J. Geotech. Geoenviron. Eng., ASCE, 137(7), 717-729. Zhou, H. and Randolph, M.F. 2006. Large deformation analysis of suction caisson installation in clay. Canadian Geotechnical J. 43, 1344-1357. Zhou, H. and Randolph, M.F. 2009a. Resistance of full-flow penetrometers in rate-dependent and strain-softening clay. Géotechnique 59(2), 79-86. Zhou, H. and Randolph, M.F. 2009b. Numerical investigations into cycling of full-flow penetrometers in soft clay. Géotechnique 59(10), 801-812.

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Ménard Lecture The pressuremeter test: Expanding its use Conférence Ménard L’essai pressiometrique : élargissement de son utilisation Briaud J.-L. President of ISSMGE, Professor, Texas A&M University, Zachry Dpt. of Civil Engineering, College Station, Texas, 778433136, USA ABSTRACT: The purpose of this contribution is to show how the use of the PMT can be expanded further than current practice. The topics covered in a first part include the amount of soil testing necessary to meet a reliability target, the influence of the lack of tensile resistance of soils on the PMT modulus, how to recreate the small strain early part of the curve lost by the decompressionrecompression process associated with the preparation of the PMT borehole, best practice for preparing the PMT borehole, commonly expected values of PMT parameters, the use of the PMT unload-reload modulus, and correlations with other soil parameters. The second part deals with foundation engineering and includes the use of the entire expansion curve to predict the load settlement behavior of shallow foundations, the load displacement behavior of deep foundations under horizontal loading, foundation design of very tall structures, long term creep loading, cyclic loading, and dynamic vehicle impact. Finally an attempt is made to generate preliminary soil liquefaction curves base on the normalized PMT limit pressure. RÉSUMÉ : Le but de cette contribution est de montrer comment l’utilisation du PMT peut être étendu au-delà de la pratique courante. Les sujets abordés dans une première partie comprennent la quantité de reconnaissance de sol nécessaire pour atteindre un objectif de fiabilité, l’influence de l’absence de résistance des sols à la traction sur le module du PMT, comment recréer la partie de la courbe en petites déformations perdue pendant la décompression-recompression associée à la préparation du trou de forage, les meilleures pratiques pour la préparation du trou de forage, les valeurs communes des paramètres PMT, l’utilisation du module déchargerecharge, et des corrélations avec d’autres paramètres du sol. La deuxième partie traite des travaux de fondation et les sujets suivants sont abordés: l’utilisation de la courbe d’expansion du PMT pour prédire le comportement des fondations superficielles, et le comportement des fondations profondes sous charge horizontale, la conception des fondations des structures de grande hauteur, le comportement de fluage, chargement cyclique, et chargement par impact de véhicules. Enfin, on propose des courbes préliminaires de liquéfaction du sol sur la base de la pression limite normalisée du PMT. KEYWORDS: pressuremeter, modulus, limit pressure, shallow foundations, deep foundations, retaining walls, liquefaction. 1

later with the corrected manuscript again rather depressed and telling Don, there is nothing left for me to do, everything has been done. Don smiled and told me don’t worry, there is much more to be done on the PMT; I feel that it is still true today and, in fact, it is the topic of this lecture. So this is my story on the PMT and I have been a fan of the PMT ever since.

HOW I GOT INTERESTED IN THE PMT?

The year is 1974 and I am a Master student at the University of New Brunswick, Canada working with Arvid Landva. I had learnt that the triaxial test was the reference test in the laboratory. I had also read from Terzaghi that the action was in the field. So I sat down one late afternoon and tried to invent an in situ triaxial test. I drew some complex systems with double tube samplers and the pressure applied between the two tubes on an internal membrane. It was very complicated and failed the Einstein test of optimum simplicity. I had also learnt from many months behind a drill rig that anything complicated had very little chance of success in the field so I kept searching and designing and then it dawned on me. What if I inverted the problem, drew an inside out triaxial test, and applied the pressure from inside the tube and pushed outward on the soil. And so I designed my first pressuremeter. I was very excited about my new invention and could not sleep that night. I waited anxiously to go to the library the next morning to see what I could dig on this idea. I went to the library and there it was Louis Menard 1957, Jean Kerisel as his advisor, the Master in Illinois with Ralph Peck, the development of the design rules, Sols Soils, 1963 and on and on. I came out of the library that morning, very disappointed that my idea had already been invented. After much reflection that day, I finally decided that I should be happy because it was obviously a good idea since it had received that much attention. This is how I got interested in the pressuremeter. I then went to The University of Ottawa to work with Don Shields who was connected with Francois Baguelin and Jean Francois Jezequel writing the pressuremeter book. Don gave me the manuscript in early Sept 1976 and said read this and correct any mistake. I did and came back 3 months

2

SPECIAL THANKS TO LOUIS MENARD

I met Louis Menard (Fig. 1) on 15 December 1977, one month before he died of cancer. I was a PhD student at the University of Ottawa in Canada working on my pressuremeter research with Don Shields. I was coming back home for Christmas that year and Louis Menard was kind enough to take some time from his very busy schedule to visit with me at the Techniques Louis Menard in Longjumeau near Paris. I waited for 30 minutes but finally got to meet the man who had invented the tool I was so fond of. Around 7 o’clock that day, I entered a huge deep office much like you see in castles. At the other end behind a big desk was Louis Menard waving at me to come closer and take a seat. I introduced myself and we started to talk about the pressuremeter. Very quickly, I found myself enjoying the discussion and time flew by. We talked and argued and talked again and quoted data and theory and reasoning so much so that at the end we had connected. I was mad because I promised myself that I would take notes of what Menard was saying but in the heat of the action I forgot all about it and was left with no notes and it was already 8 O’clock. This is where I got really lucky. Louis Menard asked me: “do you have any plans tonight? I said no and he said: “why don’t you stay for dinner?” Whaoh! That would be wonderful. We got up and he took his cane to walk from his office to his house which was a

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door away. The cancer was very advanced but he explained to me as we walked to the dining room that he had a slight illness but that he would take care of that in no time! This is where I got my first clue of the remarkable strength of his will power, the steely determination of Louis Menard, a trait of character which helped him win against all odds while creating some slight antagonistic situations. The dinner was a delight. Honestly, I cannot tell you what I ate but I certainly remember the stories that he told me with his wife and his children around the table. One stands out in my mind: his first encounter with Ralph Peck. He said that he entered Professor Peck’s office and Peck proceeded to explain to young Louis Menard that he would have to take a certain number of core courses to get his Master degree. So Peck walked to the small blackboard in his office and wrote a list of these 4 or 5 courses, then went back to his desk. Louis Menard got up, took the eraser and wiped the courses out and said I am not interested in these courses; however I am interested in these courses instead. Menard was indeed a very bright, very determined independent thinker. On that day of 15 December 1977 he provided me with a wonderful moment in my life, one that I will never forget.

the mean modulus measured on the soil samples has a 98% confidence level of being within + or – 20% of the true mean of the modulus? For this we recall the student t distribution. Consider a large population (the big cube) of modulus E which is normally distributed with a mean μp and a standard deviation σp. Then consider a group of n randomly selected values of the modulus (E1, E2, E3, …, En) from the population (results of the site investigation and testing). The mean modulus value of the group E1, …, En, is μg and the standard deviation is σg. Let’s create many such groups of n modulus values (many options of where to drill and where to test), each time randomly selecting n values from the larger population of modulus and calculating the mean modulus μg of the group. In this fashion we can create a distribution of the means μg. It can be shown that the distribution of the means μg has a mean μμg equal to μp and a standard deviation σμg equal to σp/n0.5. If we form the normalized variable t:

t

g   p

g / n

(1)

then the distribution of t is the student t distribution for n degrees of freedom: t(n). The t distribution is more scattered than the normal distribution of E, depends on the number n of modulus values collected in each group, and tends towards the normal distribution when n becomes large (Fig. 2).

Figure 1. Louis Menard (courtesy of Michel Gambin and Kenji Mori) 3

INTRODUCTION

Figure 2. The student t distribution

There are many different types of pressuremeter devices and many ways to insert the pressuremeter probe in to the ground. This paper is limited to the preboring pressuremeter also called Menard pressuremeter where a borehole is drilled, the drilling tool is removed, and the probe is lowered in the open hole. The probe diameter is in the range of 50 to 75 mm and the length of the inflatable part of the probe in the range of 0.3 to 0.6 m. The paper starts with a general observation regarding site investigations, then deals with many aspects of the pressuremeter practice including the device itself, the installation, the test, the parameters that can be obtained, and their use in foundation engineering. In each topic, new contributions are made to expand the use of the PMT. 4

HOW MANY BORINGS ARE ENOUGH?

What percentage of the total soil volume involved in the soil response should be tested during the geotechnical investigation. This depends on many factors including the goal of the investigation. This goal may be that there is a high probability that the predictions will be within a target tolerance. As an example of calculations, assume that the block of soil which will be loaded by the structure is a cube 10 x 10 x 10 m in size. Further assume that the goal is to predict the elastic settlement of the structure with a precision of + or – 20% and that the soil cube has a modulus with a coefficient of variation equal to 0.3. The question is: what percentage of the total volume of soil must be tested to have a 98% probability that the predicted settlement will be within + or - 20% of the true settlement (i.e.: measured)? Since in this case the modulus is linearly proportional to the settlement, the question can be rephrased to read: what percentage of the soil volume must be tested so that

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The properties of the student t distribution together with Eq.1 allow us to write:

 g  g   P  g  t  p   g  t 1       ,n1 n ,n1 n  2 2    

(2)

Where t(α/2,n-1) is the value of t for n-1 degrees off freedom and a value of α/2, α is the area under the t distribution for values larger than t (Fig. 3). Eq.2 expresses that there is a (1-α) degree of confidence that the value of μp is between the values expressed in the parenthesis. For our example, we need to determine the number n of modulus values in the group (number of samples to be collected and tested during the site investigation) which will lead to a high probability P that the predicted modulus (μg) will be within a target tolerance ∆ from the true mean modulus of the population (μp). Therefore we wish to find the value of n which will satisfy the probability equation:

P  g (1  )   p  g (1  )   Ptarget

Figure 3. Definition of the parameter α.

(3)

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That is to say we have a Ptarget % degree of confidence that μp lies in the range μg(1+or-∆). We can rewrite Eq.3 as

   p  g P        Ptarget   g  

such anomalies, the amount of soil volume to test would increase. If we use the same approach for different volumes we can generate the number of borings necessary to meet the criterion of 98% confidence of predicting within + or – 20% for a soil with a coefficient of variation equal to 0.3. Fig. 5 shows the number of borings required as a function of the soil volume involved in the response to the loading. The estimated line for current practice is plotted on the same graph (based on the author’s experience) indicating that current practice does not meet the criterion established. Note that the discrepancy increases with the size of the project. Indeed the ratio between the required number of borings Nr and the current number of borings Nc increases with the size of the imprint.

(4)

If the coefficient of variation of the population is δ, then we assume that the coefficient of variation of the group is also δ.

 

 p g   p g

(5)

Combining Eq.2, 4, and 5 we get.

g

    or n    t    g  t , n 1 n     2 ,n 1  2 2

2

(6)

Eq. 6 is solved by iteration since n influences the value of t. Student t distribution solvers are available on the internet. The number n represents the number of soil samples to be tested in order to obtain the value of the modulus within plus or minus ∆% from the exact answer with a Ptarget probability of success. If we assume that a triaxial test sample to obtain a modulus value has a volume of 10-3 m3, then the number n of samples gives the volume of soil that must be drilled during the investigation to satisfy the criterion. The percent volume tested becomes

Vs n  10 3  Vt Vt

(7)

In our example the initial volume was 1000 m3, so we can calculate what percentage of the soil volume should be tested. Fig. 4 gives the results and indicates that in order to be 98% sure that the answer will be within plus or minus 20% from the true value, the amount of sampling is 0.001 percent of the total volume.

Figure 5. Comparison of number of borings in current practice and number of borings required for a precision of + or - 20% with a 98% degree of confidence for a soil parameter coefficient of variation of 0.3. 5

WHAT CAN BE IMPROVED ABOUT THE PMT EQUIPMENT?

Only a few things, I think. We are at the point of maturity in this area. If anything, we need to be able to run controlled stress tests or control strain tests equally well. Controlling strain or volume has the advantage of not having to guess at the limit pressure to decide on the pressure steps. Controlling pressure has the advantage of not having to wait for a long time if the hole is too big. The devices which control stress require compressed gas bottles which can be dangerous. Control volume devices are safer in that respect and still allow control stress tests. Most civil engineering structures apply stress control steps. With regard to the issue of the three cells versus mono-cell probes, it has been shown (Briaud, 1992) that for probes with a length to diameter ratio longer than 6, the difference between the expansion of the mono-cell and the expansion of an infinitely long cylinder for an elastic soil are within 5 % of each other. Therefore as long as the probe has a length to diameter ratio of 6 or more, there is no need for three cells in a pressuremeter probe. The diameter of the probe has an impact on the quality of the test for the following reason. The thickness of the ring of disturbed soil created by the carving or washing process during drilling is approximately constant regardless of the diameter of the drill bit. As such, the larger the pressuremeter diameter is, the less influence this disturbed zone will have on the pressuremeter curve. Therefore, it is best to increase the diameter of the pressuremeter probe. A larger diameter will also have a positive impact on the reliability of the borehole diameter as it is much easier to drill a well calibrated 150mm diameter hole than a 50mm diameter hole. Using lightweight yet rugged 150 mm diameter, 1 m long PMT probes will improve PMT test quality.

Figure 4. Required volume of soil to be tested as a percent of the total volume involved in the soil response to predict a soil property with a 98% confidence level and within a percent error for given coefficients of variation of the soil property. Consider now an 8 story building which is 40 by 40 m at its base. The volume of soil involved in the response of the building to loading is at least 40 by 40 by 40 m or 64000 m3. The required sampling is 0.001% or 0.64 m3 which corresponds to 640 triaxial tests. Further assuming that we will drill 40 m deep borings allowing us to conduct 20 triaxial tests per boring, this would require some 32 borings. In practice, we would typically drill 4 or 5 borings for such a building. This shows that we do not test the soil enough in our current soil investigations to meet the set criterion. Note that the assumptions made in the student t distribution calculation include the assumption that the soil is uniformly variable. In other words, there are no heterogeneity trends or anomalies in the soil mass. If there were

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6

MAKING A QUALITY BOREHOLE IS THE MOST IMPORTANT STEP

 r  

DRILLING FOR SAMPLING Fast rotation to get to the sampling depth faster Don’t care about borehole walls left behind the bit Care about undisturbed soil in front of the bit

 z

E 

z

z

(9)

Ez



E z  z  E z

(12)

Er zr  E z rz

(13)

Here it is assumed that a compression modulus E+ acts in the radial and vertical direction and a much reduced tension modulus E- acts in the hoop direction. (14) E E E z r

E  E 

(15)

+

Where E is the modulus of the soil when tested in compression and E- is the modulus of the soil when tested in tension. The problem is further simplified by assuming that (16)  rz   zr   1

Stop at sampling depth Clean borehole by running bit with fast mud flow up and down in open hole; avoids unwanted cuttings in sampling tube Don’t care about borehole diameter

Care about borehole diameter



(10) Er E Ez Where εr, εθ, εz are the normal strains in the r, θ, and z directions, σr, σθ, σz are the normal stresses in the r, θ, and z directions, Er, Eθ, Ez are the modulus in the r, θ, and z directions, and νθr, νrθ, νzr, νrz, νzθ, νθz are the Poisson’s ratios. Because of the symmetry rules, the following equations must also be satisfied (11) Er  r  E r

Table 1. Differences between drilling for PMT testing and drilling for soil sampling

Do not clean the borehole by running the bit up and down in the open hole; this will increase the hole diameter

Er



  rz r   z z 

This is the most important and the most difficult step in a quality pressuremeter test. Much has been tried and written on the best way to prepare the hole. Special training is required for drillers to prepare a good PMT borehole as drilling for PMT testing is very different and almost opposite to drilling for soil sampling (Table 1). Table 2 gives some general recommendations to obtain a quality borehole with wet rotary drilling which I would recommend in most cases.

DRILLING FOR PMT TESTING Slow rotation to minimize enlargement of borehole diameter Care about undisturbed borehole walls left behind the bit Don’t care about soil in front of the bit Advance borehole beyond testing depth for cuttings to settle in

r

 z   r   2  z   r   3

(17) (18)

The plane strain condition of the cylindrical deformation gives (19) z  0 The definition of the strains is, in small strain theory

du dr u   r

Table 2. Recommendations for a quality PMT borehole by the wet rotary method.

r 

Diameter of drilling bit should be equal to the diameter of the probe Three wing bit for silts and clays (carving), roller bit for sands and gravels (washing) Diameter of rods should be small enough to allow cuttings to go by Slow rotation of the drill (60 rpm) Slow mud circulation to minimize erosion Drill 1 m past the testing depth for cuttings to settle One pass down and one withdrawal (no cleaning of the hole) One test at a time

7

d r  r     0 dr r

E

(22)

Using Eq. 8 to 22 leads to the governing differential equation where the displacement u is the variable. The boundary conditions are a displacement equal to zero for an infinite radius and a pressure equal to the imposed pressure at the cavity wall. The solution is a bit cumbersome: 1 (23)  u  ro   s12  ( s21  s12 )  ( s21  s12 ) 2  4 s11 s22  o 

7.1 PMT Modulus and tension in the hoop direction A number of parameters are obtained from the PMT. One of the most useful is the PMT modulus Eo from first loading This modulus is calculated by using the theory of elasticity. One of the assumptions in elasticity is that the soil has the same modulus in compression and in tension. This may be true to some extent for clays but unlikely true for sands. When the PMT probe expands, the radial stress increases and the hoop stress decreases to the point where it can reach tension. In elasticity, the increase in radial stress is equal to the decrease in hoop stress, so if the pressure in the PMT probe is 500 kPa, the hoop stress at the borehole wall is -500 kPa (neglecting the at rest pressure). The soil is unlikely to be able to resist such tension and using elasticity theory in this case is flawed. The following derivation shows the influence of having a much weaker modulus in tension than in compression. The general orthotropic elastic equations are

Er

(21)

Now the equilibrium equation gives

THE PMT PARAMETERS

    r r   r   zr z

(20)

2

 r



o

Where s11, s22, s12, s21 are defined as follows

s11 

E  1  22 

1  2

2 2

 1  1 1 

(24)

s12 

E  2 1  2 22  1 

(25)

s21 

E  2 1  2 22  1 

(26)

s22 

E  1  1 

1  2

2 2

 1 

(27)

Eq. 23 is to be compared with the equation for the isotropic solution which is

 E u

(8)

 ro   o  o  1    ro

Ez

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(28)

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Consider the case where the ratio E+/E- = 10, ν1 = ν3 = 0.33, then ν2 equal to 0.033. Then Eq.23 and Eq. 28 give respectively: u (29)  ro  0.309 E  o ro

expansion is defined as the radius at which εθ and Δσr are 1/10th of the value at the cavity wall, that radius of influence is 100.5Ro = 3.16Ro. Within this radius of influence, the average strain εθ can be calculated as follows 2 3.16 Ro  R 1 (33) o o uo   av  dR 0.316 o 2 (30)   ro  0.752 Eo 3.16 Ro  Ro  Ro R  ro where εθav is the average hoop strain within the radius of Therefore, E+ = 2.43Eo (31) influence of the pressuremeter test, εθo is the hoop strain at the + This can be repeated for different values of E /E to obtain Fig. wall of the cavity, Ro is the initial radius of the cavity, and R is 6. The inverse of the modulus ratio is consistent with the values the radial distance in the soil. The modulus was mentioned as recommended by Menard for the α values in settlement analysis being associated with a strain level at the cavity wall εθo as shown in Fig.6. This observation about the tension in the typically in the range of 2 to 6%; this means that the average hoop direction also impacts PMT tests in hard soils and rock strain εθav will be 0.6 to 2%. For the two Texas sites mentioned which are sound enough to exhibit significant tensile strength. above, the average strain would be close to 1% (3.53% x 0.316). In this case, the PMT curve shows a break in the expansion Note that this range of strain is consistent with the strain level curve (Fig. 7) at a pressure p where the hard soil or rock breaks associated with foundation engineering but is much higher than in tension. This pressure is such that (Briaud, 1992): the range of strain associated with pavement design or (32)  t p  2 oh earthquake shaking where a very low strain modulus is used. The fact that the small strain modulus is absent from the Where σt is the soil tensile strength and σoh is the horizontal beginning of the PMT curve and that the strain range is between stress at rest before the PMT is inserted. 0.6 to 2%, is created in part by the recompression of the soil which was decompressed horizontally by the drilling process. This recompression makes the small strain part of the stress strain curve disappear as shown in the PMT test on Fig. 8. In this test, an unload-reload loop was performed by decreasing the pressure to zero and increasing it again to simulate a first expansion curve. Then a second unload-reload loop was performed over a much smaller pressure range. This test shows that the recompression modulus varies tremendously depending on the extent of the unloading. This test also shows that the low strain information is lost in the decompression and recompression loading process. Can we find a way to recreate the early part of the PMT curve from the information gathered during the test. Figure 6. Correction of PMT modulus for low tension soils

CONCEPT

ACTUAL TEST 1400

Figure 7. Tensile strength from PMT test

1200 1000

P (kPa)

7.2 PMT first load modulus The PMT first load modulus Eo also called the Menard modulus is obtained from the initial straight line part of the PMT curve. This straight line exists over a range of relative increase in cavity radius which varies from one soil to another but is typically in the range of 2 to 6 % relative increase in cavity radius. At two sites in Texas, one in stiff clay the other in dense sand, the average range of 15 PMT tests was 3.47% for the clay site and 3.59% for the sand site. This refers to the value of ΔR/Ro at the cavity wall. The average radial strain in the soil mass involved in the response to the cylindrical cavity expansion is much smaller and averages 0.316 ΔR/Ro as shown in the following. The hoop strain εθ and the increase in radial stress Δσr decrease away from the wall of the cavity at a rate of 1/R2 where R is the radial distance into the soil mass (Baguelin et al., 1978). If the radius of influence of the pressuremeter

800 600 400 200 0 0.00

0.04

0.08 0.12 dR/R0

0.16

0.20

Figure 8. PMT stress strain curve with unload reload loops 7.3 PMT modulus at small strain A soil modulus depends on several factors (Briaud, 2013) one of which is the strain level. The PMT curve is a stress strain curve where the stress is the radial stress σr (measured pressure

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in the PMT) and the strain is the hoop strain εθ (relative increase in cavity radius). It is therefore possible to define a secant modulus as a function of strain from the PMT curve (Fig. 9).

a. REZEROED PMT CURVE

b. HYPERBOLIC CURVE FITTING

Figure 9. PMT stress strain curve and secant modulus It can be shown in elasticity that the shear modulus is given by: 1  ro (34) G 2  o If we call Go the shear modulus associated with the straight portion of the curve, we can normalize the modulus at any strain with respect to Go. We calculate the secant shear modulus G1, G2, G3 and so on corresponding to points 1, 2, and 3 on the pressuremeter curve (Fig. 9). Then we can plot the ratio G1/Go, G2/Go, G3/Go as a function of the corresponding strain εθ1, εθ2, εθ3. Note that εθ is the strain at the cavity wall but that the mean strain εθmean induced in the soil within the zone of influence is only about 32% of that value (Eq. 33). The curve linking G/Go vs. εθmean is shown on Fig. 10c and 10d. From zero strain to the strain value corresponding to the end of the straight part of the PMT curve (AB on Fig. 10a), the G/Go vs. εθmean curve is flat on Fig. 10c and 10d because within that strain range the modulus G is constant and equal to Go. In order to generate the non linear beginning of that curve (EB on Fig. 10a), it is convenient to assume a hyperbolic model as proposed by Baud et al. (2013) of the form (35)   1   2Gmax pL This equation defines a hyperbola which describes the PMT curve with the limit pressure pL as the asymptotic value and 2Gmax as the initial tangent modulus. The hyperbolic model has been shown to be very successful in describing the stress strain curve of soils (Duncan, Chang, 1970). In Eq. 35, pL is known and all the points on the PMT curve, after excluding the points on the straight line part, can be used to find the optimum value of Gmax by best fit regression. This can be done by plotting the data points as ε/σ vs. ε and fitting a straight line through the data points (Fig. 10b). Then 1/2Gmax is the ordinate at ε = 0 and 1/pL is the slope of the line.  1  (36)    2Gmax pL Then Eq. 35 gives the complete curve. This technique was used at two sites, a stiff clay site near Houston, Texas, and a medium dense sand site in Corpus Christi, Texas. Example results are presented in Fig. 11 which shows that the data fits well with a hyperbolic equation. For these two sites, the average ratio Gmax/Go was 1.75 for the stiff clay and 1.27 for the dense sand.

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c . NORMALIZED SECANT SHEAR MODULUS VS STRAIN

d . NORMALIZED SECANT SHEAR MODULUS VS LOG OF STRAIN

Figure 10. Normalized secant shear modulus vs. strain Estimates of Gmax were calculated independently by using correlations proposed by Seed et al. (1986) based on SPT blow count for sand, Rix and Stokoe (1991) based on CPT point resistance for sand, and Mayne and Rix (1993) based on CPT point resistance and void ratio for clays. These estimates of Gmax were consistently much higher than the values obtained by the hyperbolic extension of the PMT curve; 25 times larger for the stiff clay and 44 times larger for the dense sand. This indicates that this hyperbolic fit to the PMT curve does not lead to accurate very small strain moduli.

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e. PMT CURVE – DENSE SAND

a. PMT CURVE – STIFF CLAY

f. HYPERBOLIC CURVE FITTING b. HYPERBOLIC CURVE FITTING

g. NORMALIZED SECANT SHEAR MODULUS VS STRAIN c. NORMALIZED SECANT SHEAR MODULUS VS STRAIN

h. NORMALIZED SHEAR MODULUS VS LOG STRAIN d. NORMALIZED SHEAR MODULUS VS LOG STRAIN

Figure 11. Examples of hyperbolic extension of the PMT curve (stiff clay, dense sand)

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7.4 PMT modulus long term creep, and cyclic loading It is relatively easy to maintain the pressure constant during a PMT test while recording the increase in radius of the cavity (Fig. 12). A pressure holding step of 10 minutes is not very time consuming and can lead to very valuable information if the structure will be subjected to long term loading (e.g.: building, retaining wall). The pressure held for 10 minutes should be higher than 0.2pL because below that threshold the influence of the decompression-recompression effect and the disturbance effect is more pronounced (Briaud, 1992). The evolution of the secant modulus Et during the pressure holding test is well described by the following model:

LOAD (MN) 0

2

4

6

8

10

12

0 -20

SETTLEMENT (mm)

t Et  Eto    to 

a. FOOTING LOAD-SETTLEMENT CURVE

n

(37)

-40 -60 -80 -100 -120 -140 -160

Where t is the time after the start of the pressure holding step, to is a reference time after the start of the pressure holding step usually taken as 1 minute, Et and Eto are the secant modulus corresponding to t and to respectively, and n is the creep exponent. The value of n is obtained as the slope of the plot of log Et/Eto vs. log t/to. The creep exponent n increases with the stress applied over strength ratio and depends on the soil type and stress history. It has been found in the range of 0.01 to 0.03 for sands and in the rnage of 0.03 to 0.08 for clays (Briaud, 1992). For clays, the lower values are for overconsolidated clays while the higher values are for very soft clays. Measurements on large scale spread footings on an unsaturated silty sand (Briaud, Gibbens, 1999) demonstrated that the power law model works very well (Fig. 13) because the log settlement vs. log time curve was remarkably linear. These experiments also indicated that n increases with the load level but is significantly reduced by unload reload cycles. PMT tests with creep steps were performed next to the footings (Fig. 13c and 13d); the parallel between the footing and the PMT is striking.

b. FOOTING SETTLEMENT VS TIME CURVE LOG DISPLACEMENT LOG10 (S/S1)

0.06

0.05

0.04

6.23 MN 7.12 MN 8.01 MN 8.9 MN 9.79 MN 10.24 MN

0.03

0.02

0.01

0.00 0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6

LOG TIME, LOG10(t/t1)

c. PMT STRESS VS STRAIN CURVE

a. CREEP TEST

d. PMT MODULUS VS TIME CURVE

b. CYCLIC TEST

Figure 13. Creep response of a 3m by 3m spread footing and a PMT test (Briaud, Gibbens, 1999, Jeanjean, 1995). Similarly, one can conduct cyclic loading during the PMT test. A series of 10 cycles is not very time consuming and can lead to very valuable information if the structure will be

Figure 12. Creep and cyclic PMT test

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subjected to significant repeated loading (e.g.: large wave loading). The evolution of the secant modulus EN to the top of cycle N is well described by the following model (38) EN  E1 N  m

a. PILE LOAD-DISPLACEMENT CURVE

Where N is the number of cycle using number 1 as the first loading cycle, EN the secant modulus to the top of the Nth cycle, E1 the secant modulus to the top of the first cycle (first time that the pressure is decreased), and m is the cyclic exponent. The value of m is obtained as the slope of the plot of log EN/E1 vs. log N. Fig. 14 shows a parallel example of a pile subjected to cyclic horizontal loading and a cyclic PMT test. As can be seen the power law model of Eq.38 describes the evolution of the deformation with the number of cycles (straight line on log-log scales) very well and the parallel between the pile and the PMT is striking. 7.5 PMT unload-reload modulus The unload reload modulus Er is obtained by performing an unload reload loop during the PMT test. The main problem with Er is that, unlike Eo, it is not precisely defined. Indeed it depends on the strain amplitude over which the loop is performed and to a lesser extent on the stress level at which the loop is performed. As such, Er varies widely from one user to another and cannot be relied upon for standard calculations unless the strain amplitude and stress level have been selected to match the problem at hand. In my practice, I perform an unload reload loop at the end of the linear phase and unload until the pressure has reached one half of the peak pressure. This has the advantage of being consistent but does not necessarily correspond to a consistent strain amplitude from one test to the next. I would strongly discourage the use of the reload modulus because it is not a standard modulus. Instead I would recommend the use of a hyperbolic extension of the PMT curve to find the modulus at the right strain level. 7.6 The yield pressure py. The yield pressure py is found at the end of the straight line corresponding to the PMT modulus. Up to py, the amount of creep is reasonably small but becomes much larger beyond that. In geotechnical engineering it is always desirable to apply pressures on the soil below the value of py. Typically py is 0.5 pL for clays and 0.33 pL for sands. Therefore, at working loads, it is advisable to keep the pressure under foundations at most equal to 0.5 pL in clays and 0.33 pL in sands to limit creep deformations. 7.7 Correlations between PMT parameters and other soil parameters Correlations based on 426 PMT tests performed at 36 sites in sand and 44 sites in clay along with other measured soil parameters were presented by Briaud (1992). These correlations exhibit significant scatter and should be used with caution. Nevertheless they are very useful in preliminary calculations and for estimate purposes. Table 3 gives the range of expected PMT limit pressure and modulus in various soils while Tables 4 and 5 give the correlations.

b. PILE STIFFNESS VS NUMBER OF CYCLES CURVE

c. PMT STRESS STRAIN CURVE

d. PMT MODULUS VS NUMBER OF CYCLES CURVE

Table 3. Expected values of Eo and PL in soils CLAY Soil strength p*L(kPa) E0 (MPa) Soil strength p*L(kPa) E0(MPa)

Soft

Medium

0–200 0 – 2.5

200–400 2.5 - 5.0

Stiff 400–800 5.0 - 12 SAND

Very Stiff

Hard

800-1600 12 - 25

>1600 > 25

Loose

Compact

Dense

Very Dense

0 – 500 0 – 3.5

500 - 1500 3.5 - 12

1500-2500 12 – 22.5

> 2500 > 22.5

Figure 14. Cyclic response of a laterally loaded pile A and a PMT test (Little, Briaud, 1988).

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

Table 4. Correlations for Sand (Column A = Number in Table x Row B) B A E0 (kPa) ER (kPa) p*L (kPa) qc (kPa) fs (kPa) N (bpf)

Column A = number in table x row B E0 ER p*L qc fs (kPa) (kPa) (kPa) (kPa) (kPa) 1

0.125

8

1.15

N (bpf)

57.5

383

8

Strength parameter

Clay

Sand

PMT pL(kPa)

1.25

1.7

CPT qc(kPa)

0.3

0.2

SPT N(bpf)*

60

75

* Ultimate bearing capacity pu in kPa. 8

1

64

6.25

312.5

2174

0.125

0.0156

1

0.11

5.5

47.9

0.87

0.16

9

1

50

436

0.0174

0.0032

0.182

0.02

1

9.58

0.0026

0.00046

0.021

0.0021

0.104

1

Table 5. Correlations for Clay (Column A = Number in Table x Row B) B A E0 (kPa) ER (kPa) p*L (kPa) qc (kPa) fs (kPa) su (kPa) N (bpf)

Table 6. Bearing capacity factors k for in situ tests

E0 (kPa)

Column A = number in table x row B ER p*L qc fs su (kPa) (kPa) (kPa) (kPa) (kPa)

N (bpf)

1

0.278

14

2.5

56

100

667

3.6

1

50

13

260

300

2000

0.071

0.02

1

0.2

4

7.5

50

0.40

0.077

5

1

20

27

180

0.25

0.05

1

1.6

10.7

0.133

0.037

1

6.7

0.02

0.005 6

0.14

1

0.079 0.010 0.001 5

0.003 8 0.003 3 0.000 5

0.62 5 0.09 1

SHALLOW FOUNDATIONS

8.1 Ultimate bearing capacity The general bearing capacity equation for a strip footing is:

1 pu  c ' Nc   BN   DNq 2

(39)

Where pu is the ultimate bearing pressure, c’ the effective stress cohesion intercept, γ the effective unit weight of the soil, Nc, Nγ, and Nq bearing capacity factors depending on the friction angle φ’. The assumptions made to develop this equation include that the unit weight and the friction angle of the soil are constant. Therefore the strength profile of the soil is linearly increasing with depth. For strength profiles which do not increase linearly with depth, this equation does not work and can severely overestimate the value of pu. However equations of the following form always take into account the proper soil strength: (40) p ks D u Where k is a bearing capacity factor, s is a strength parameter for the soil, γ is the unit weight of the soil, and D is the depth of embedment. The parameter s can be the PMT limit pressure pL, the CPT point resistance qc, or the SPT blow count N. Table 6 gives the values of k for various soils and various tests in the case of a horizontal square foundation on horizontal flat ground under axial vertical load.

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8.2 Load settlement curve method for footings on sand The typical approach in the design of shallow foundations is to calculate the ultimate bearing capacity pu, reduce that pressure to a safe pressure psafe by applying a combined load and resistance factor, use that safe pressure to calculate the corresponding settlement, compare that settlement to the allowable settlement, and adjust the footing size until both the ultimate limit state and the serviceability limit state are satisfied. In other words the design of shallow foundations defines two points on the load settlement curve: one for the ultimate load and one for the service load. It would be more convenient if the entire load settlement curve could be generated. Then the engineer could decide where, on that curve, the foundation should operate. This was the incentive to develop the load settlement curve method (Briaud, 2007). Five very large spread footings on sand up to 3m x 3m in size were loaded up to 12 MN at the Texas A&M University National Geotechnical Experimentation Site (Fig. 15a). Inclinometer casings were installed at the edge of the footings as part of the instrumentation. They were read at various loads during the test and indicated that the soil was deforming in a barrel like shape (Fig. 15b). This is the reason why the pressuremeter curve was thought to be a good candidate to generate the load settlement curve for the footing. Note that, during these tests, the inclinometers never showed the type of wedge failure assumed in the general bearing capacity equation. It is reasonned that the footings were not pushed to sufficient penetration to generate this type of failure mechanism. The transformation required a correspondence principle between a point on the pressuremeter curve and a point on the footing load settlement curve (Fig. 16). This correspondence was established on the basis of two equations: the first one would satisfy average strain compatibility between the two loading processes and the second one would transform the PMT pressure into the footing pressure for corresponding average strains. These equations are: s R (41)  0.24 B Ro

p f  f L / B f e f f  , d  p p

(42)

Where s if the footing settlement, B the footing width, ∆R/Ro the relative increase in cavity radius in the PMT test, pf the average pressure under the footing for a settlement s, fL/B, fe, fδ, fβ,d the correction factors to take into account the shape of the footing, the eccentricity of the load, the inclination of the load, and the proximity of a slope respectively, Γ a function of s/B, and pp the pressuremeter pressure corresponding to ∆R/Ro. The Γ function was originally obtained from the large scale footing load tests on sand at Texas A&M University (Jeanjean, 1995, Briaud, 2007) and then supplemented with other load tests. This led to the data shown on Fig. 17. Using all the curves (Fig. 17a), a mean and a design Γ function were obtained (Fig. 17b). The design Γ function curve is the mean Γ function curve minus one standard deviation. The f correction factors have been determined through a series of numerical simulations previously calibrated against the large scale loading tests (Hossain, 1996, Briaud, 2007). Their expressions are as follows

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a. LOAD TEST SET UP Near 2/1 slope

8.5 m Settlement Beam

LVDT

Jack

Telltates 2B

Inclinometer casings

10.7 m

15 m

B Dywidag bars only No concrete

0.15

(49)

Where B is the width of the footing, L its length, e the load eccentricity, δ the load inclination in degrees, and d the horizontal distance from the slope-side edge of the footing to the slope crest. The shape of the Γ function indicates that at larger strain levels the need to correct the PMT curve is minimal. Indeed for s/B larger than 0.03, the mean value of Γ is constant and equal to about 1.5. For values of s/B smaller than 0.03, there is a need to correct the value of the PMT pressure because of a lack of curvature on the PMT curve compared to the curvature on the footing load settlement curve.

Load cell

B 0.5B SAND

 d  f  ,d 0.7 1    B

Q LENGTH = L δ

d Drilled shaft (Concrete+Bars)

7.6 m

β

Steel plates

2.7 m

2.7 m

PRESSUREMETER-LIKE LATERAL DEFORMATION FROM INCLINOMETER

PRESSURE on WALL

b.

Pressuremeter Test

RELATIVE INCREASE IN CAVITY RADIUS

e B

Foundation D Sand Pressuremeter Test

?

LOAD SETTLEMENT

CLAY SHALE

Foundation Behavior

Figure 16. Transformation of the pressuremeter curve into the footing load settlement curve a. Γ FUNCTION: ALL DATA

Figure 15. Analogy between the soil deformation under a shallow foundation and around a pressuremeter expansion test Shape

0.8  0.2 f L /B

Eccentricity

f e  1  0.33

B L

(43)

e B

center

(44)

edge

(45)

center

(46)

edge

(47)

b. Γ FUNCTION: DESIGN RECOMMENDATIONS

0.5

Eccentricity

Inclination

Inclination

Near 3/1 slope

e fe  1    B 2  

f  1     90 

   f  1     360 

0.5

 d f  ,d 0.8 1     B

0.1

Figure 17. The Γ function for the load settlement curve method (Briaud 2013)

(48)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

8.3 Load settlement curve method for footings on stiff clay The load settlement curve method developed for sand was extended to stiff clay by using some footing load tests and parallel PMT tests. O’Neill and Sheikh (1985) load tested a 2.4 m diameter bored and under-reamed pile in Houston (Fig. 18a). The pile was 2.4 m deep (relative embedment depth D/B = 1) and the shaft friction was disabled by a casing. The soil was a stiff clay with an undrained shear strength of about 100 kPa. The load was increased in equal load steps and the resulting load settlement curve is shown in Fig. 18b. At failure, the average pressure under the footing was 680 kPa as measured by pressure cells on the bottom of the under-ream. Briaud et al. (1985) performed pressuremeter tests at the same site around the same time. The PMT test was carried out at a depth of 3.6 m or half a diameter below the bottom of the footing; this PMT curve (Fig.19a) was used to generate the Γ function for that stiff clay (Fig. 19b). As can be seen, the curve for that stiff clay is very close to the recommended mean curve for sand. Load tests on stiff clay using a 0.76m diameter plate at a depth of 1.52m (Tand, 2013) were also analyzed together with parallel PMT tests (Briaud, 1985) and gave the other Γ functions on Fig.19b. These tests on stiff clay give an indication that the design Γ function of Fig. 17b is equally applicable to sands and stiff clays. Note that the load settlement curve method gives the response of the footing as measured in load tests. These load tests are carried out in a few hours; if the loading time is very different (one week or more or one second or less), the time effect must be considered separately (Section 7.4).

a. PMT CURVE

b. THE Γ FUNCTION

a. LOAD TEST SET UP

Figure 19. Pressuremeter test (Briaud et al, 1985) and Γ function for stiff clay b. LOAD TEST RESULTS

9

Figure 18. Large scale footing load test in stiff clay in Houston (O’Neill, Sheikh, 1985)

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DEEP FOUNDATIONS UNDER VERTICAL LOADS

The rules developed by the French administration (Fascicule 62, 1993) for calculating the vertical capacity of piles are based on a very impressive database of load tests carried out by Bustamante and Gianeselli and the Laboratoires des Ponts et Chaussees from about 1975 to 1995. These rules were recently updated (NF P94-262, 2012) and represent one of the most complete and detailed axial capacity methodology in existence. These rules should be followed closely as there is no viable alternative for the PMT. One area of deep foundations where the pressuremeter has seen some expanded use is the foundation design of very tall buildings such as the 452 m high Petronas Towers in Kuala Lumpur, Malaysia (Baker, 2010), the 828 m high Burj Khalifa in Dubai, UAE (Poulos 2009), the planned 1000 m high Nakheel Tower in Dubai, UAE (Haberfield, Paul, 2010), and the planned 1000m+ Kingdom Tower in Jeddah, Saudi Arabia (Poeppel, 2013). It is also seeing increased use for very large foundations such as the I10/I19 freeway interchange in Tucson, USA (Samtani, Liu, 2005). The use of the PMT for very tall buildings started with the work of Clyde Baker between 1965 and 1985 (Baker, 2005) for the Chicago high-rises where the use of the pressuremeter in the glacial till allowed Clyde Baker to increase the allowable pressure at the bottom of bored piles from 1.4 MPa to 2.4 MPa. The 1.4 MPa value was based on unconfined compression tests; the use of the pressuremeter along with observations led to using the 2.4 MPa value as confidence was gained.

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In making settlement calculations for such structures, some use the rules proposed by Menard and some use the elastic equations often with an unload-reload modulus. Those who use the Menard rules, use α values based on local experience and influenced by the ratio between the unload-reload modulus Er and the first load modulus Eo. While the value of the ratio Eo/Er varies within a range somewhat similar to the range of α values, it is not clear why one should be related to the other. The ratio Eo/Er is influenced by the development of plastic deformation around the probe while the value of α is argued to be related to the combination of lack of strength in tension (hoop direction as shown in Section 7.1) and recompression process through an S shape curve (Fig. 8). Those who use the elastic equation together with an unload-reload modulus face the problem that the unload reload modulus is ill defined and depends in particular on the extent of the unloading and the stress level at which the unloading takes place. The case of the foundation of the tallest tower on Earth, the 828m high Burj Khalifa in Dubai, UAE, is studied further to investigate the issue of the first load modulus and the reload modulus (Poulos, 2009). The Burj Khalifa weighs approximately 5000MN and has a foundation imprint of about 3300m2. The foundation is a combined pile raft 3.5 m thick founded at a depth of about 10 m below ground level on 1.5 m diameter bored piles extending some 50 m below the raft. To predict the settlement of the tower, a number of methods were used including numerical simulations. For these simulations a modulus profile was selected from all soil data available including 40 PMT tests. The PMT first load modulus profile is shown in Fig. 20 along with the selected design profile as input for settlement calculations by numerical simulations. As can be seen the design profile splits the PMT first load modulus profile with some conservatism. The settlement of the tower was predicted to be 77mm; it was measured during construction and reached 45 mm at the end of construction (Fig. 21). The reasonable comparison between measured and predicted settlement for this major case history gives an indication that it is appropriate to use the PMT first load modulus for settlement estimates.

Figure 21. Measured and predicted settlement of the Burj Khalifa, Dubai, UAE (after Poulos, 2009) 10 DEEP FOUNDATIONS UNDER HORIZONTAL LOADS 10.1 Single pile behavior For vertically loaded piles, it is common to calculate the ultimate capacity of the pile due to soil failure and then the settlement at working load. For horizontally loaded piles, an ultimate load due to soil failure is not usually calculated. Briaud (1997) proposed an equation to calculate the ultimate horizontal load due to soil failure for a horizontally loaded pile.      D v  4  lo for L  3lo    (50) L  for L  lo 3 D v 3 H ou  pL BD v  4 1/4   4E I  lo   p    K    K 2.3Eo  Where Hou is the horizontal load corresponding to a horizontal displacement equal to 0.1B, B the pile diameter, pL the PMT limit pressure, Dv the depth corresponding to zero shear force and maximum bending moment, lo the transfer length, L the pile length, Ep the modulus of the pile material, I the moment of inertia of the pile around the bending axis, K the soil stiffness, and Eo the PMT first load modulus. In order to expand that solution to create the entire load displacement curve for horizontally loaded piles, it is proposed to first use a strain compatibility equation such that the relative displacement to reach the ultimate load on the pile (y/B = 0.1) corresponds to the relative PMT expansion at the limit pressure (∆R/Ro = 0.41). y R (51)  0.24 B Ro Then the load on the pile can be transformed into a pressure within the most contributing zone as

p pile 

Figure 20. First load PMT modulus profile and selected design modulus values for the Burj Khalifa, Dubai, UAE (after Poulos, 2009)

119

Ho BDv

(52)

The Γ value is the ratio of the pressure on the pile divided by the pressure on the PMT for a corresponding set of values of y/B and ∆R/Ro which satisfy Eq. 51. That way and point by point, the Γ function can be generated as a function of y/B or 0.24∆R/Ro. This approach is consistent with the approach taken for the load settlement curve method for shallow foundations. This was done for 5 piles including driven and bored piles as well as sand and clay soils. The piles are described in Briaud (1997) and in Briaud et al. (1985). They ranged from 0.3 to 1.2 m in diameter and from 6 to 36 m in length. In each case, the pile dimensions were known, the load displacement curve was

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

known and the PMT curves were measured at various depths within the depth Dv. An average PMT curve was created within Dv if more than one test was available. The Γ functions obtained from these load tests and parallel PMT tests are shown in Fig. 22. They have a shape similar to the one for the shallow foundations but the pile installation seems to make a difference. The driven piles lead to one class of Γ functions while the bored pile leads to a lower function. More data would help refine this first observation.

Figure 23. Plan view of a group of horizontally loaded piles.

Figure 24. Leading pile and trailing pile efficiency factors Figure 22. The Γ functions for transforming the PMT curve into a horizontal load – displacement curve for a pile. 10.2 Pile group behavior The behavior of vertically loaded pile groups is often predicted by making use of an efficiency factor of the form (53) Qg  ev nQs Where Qg is the vertical load on the group, ev the efficiency of the vertically loaded group, n the number of piles in the group, and Qs the vertical load on the single pile for the same settlement as the pile group. This approach can be extended to the problem of horizontal loading on a pile group by writing (54) H g  ehnHs Where Hg is the horizontal load on the group, eh the efficiency of the horizontally loaded group, n the number of piles in the group, and Hs the horizontal load on the single pile for the same horizontal movement as the pile group. Fig. 23 shows the plan view of a group of horizontally loaded piles. A distinction is made between the leading piles on the front row of the group and the trailing piles behind the front row. Using data by Cox et al. (1983), Briaud (2013) proposed to extend Eq. 54 to read:

elp   Hg  (nlpelp  ntpetp )Hs   nlpelp  ntp  Hs  

(55)

Where nlp and ntp are the number of leading piles and trailing piles in the group respectively, elp and etp are the efficiency factors for the leading pile and trailing pile respectively, and λ is the ratio of elp over etp. Fig. 24 and 25 give the efficiency factors as a function of the relative pile spacing based on the data by Cox et al. (1983).

120

Figure 25. Ratio of leading over trailing pile efficiency factor Eq. 52 was developed based on ultimate load observations at large horizontal displacements. The use of the same equation for all range of horizontal movements was investigated by comparing measured and predicted movements for two major pile group experiments by Brown and Reese (1985) in stiff clay and by Morrison and Reese (1986) in medium dense sand. The plan view of the group is shown in Fig.23. The piles were 0.273m in diameter, 13.1m long steel pipe piles driven in a 3 by 3 group with a spacing of 3 diameter center to center. The group was built to simulate a rigid cap condition which is most common. The clay was a stiff clay which had an undrained shear strength of about 100kPa within the top 3 m from the ground surface. The sand was a medium dense fine sand with a CPT point resistance increasing from zero at the ground surface to 3000 kPa at a depth of 2 m. Fig. 26 presents the result for the test in clay and Fig. 27 for the test in sand. In each case, the measured load-displacement curve for the single pile is presented as well as the measured curve linking the average load per pile in the group and the group displacement. The

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efficiency in Eq. 55 was calculated as follows using Fig. 24 and 25:

yg  ys

elp   0.95   Hg  9 0.82 Hs (56)  nlpelp  ntp  Hs   3 0.95  6  Hs   1.25   

Bg  Bs

1.91  2.65 0.273

(58)

The curve linking the average load per pile in the group versus group displacement was obtained by using the load versus displacement curve for the single pile and, for any given horizontal movement, multiplying the single pile movement by 2.65. That predicted curve is shown on Fig. 28 and 29 along with the curve measured by Brown and Reese for their test in clay (1985) and Morrison and Reese for their test in sand (1986) respectively. The measured single pile curve is also shown for reference.

The predicted curve describing the average horizontal load per pile in the group versus the group horizontal displacement was obtained by using the horizontal load versus horizontal displacement curve for the single pile and multiplying the single pile load by 0.82 for any given movement. The curve predicted using this approach is shown on Fig. 26 (clay) and 27 (sand) along with the measured curves.

Figure 26. Predicted by Cox efficiency factor method and measured load-displacement curve for Brown-Reese group test in clay (1985)

Figure 28. Predicted by O’Neill efficiency factor method and measured load-displacement curve for Brown-Reese group test in clay (1985)

Figure 27. Predicted by Cox efficiency factor method and measured load-displacement curve for Morrison-Reese group test in sand (1986)

Figure 29. Predicted by O’Neill efficiency factor method and measured load-displacement curve for Morrison-Reese group test in sand (1986)

O’Neill (1983) suggested that the best and simplest efficiency factor to use for the settlement of a group of vertically loaded piles was:

11 HORIZONTAL IMPACT LOADING FROM VEHICLE

sg ss



Bg

In the case of road side safety, embassy defense against terrorist trucks, ship berthing, piles are impacted horizontally. To predict the behavior of piles subjected to horizontal impact, it is possible to use 4D programs (x, y, z, t) such as LSDYNA (2006). This is expensive and time consuming. The problem can be simplified by using a P-y curve approach generalized to include the effect of time. In this case the governing differential equation is

(57)

Bs

Where ss is the settlement of the single pile under the working load Q, sg the settlement of the group under nQ, n the number of piles in the group, Bg the width of the group and Bs the width of the single pile. This efficiency factor for the Brown and Reese pile group was (Fig. 23)

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

4 y 2 y y (59) M C  Ky  0 4 z t 2 t where E (N/m2) is the modulus of elasticity of the pile, I (m4) the moment of inertia of the pile against bending around the horizontal axis perpendicular to impact, y (m) the pile horizontal displacement at a depth z and a time t, M (kg/m) the mass per unit length of pile (mass of pile Mp plus mass of associated soil Ms), C (N.s/m2) the damping of the system per unit length of pile, and K (N/m2) the soil spring stiffness per unit length of pile. Note that the soil horizontal resistance is limited to pu (kN/m2). The boundary conditions are zero moment and zero shear at the point of impact, and zero moment and zero shear at the bottom of the pile. The initial condition is the displacement of the impact node during the first time step; this displacement is equal to vo x Δt where vo is the velocity of the vehicle and Δt the time step. Other inputs include the mass and velocity of the impacting vehicle, and the parameters in Eq. 59 for the soil and the pile. The differential equation is then solved by the finite difference method and it turns out that the parameter matrix is a diagonal matrix so that no inversion is necessary. As a result the solution can be provided in a simple Excel spread sheet (Mirdamadi, 2013). Because the problem is a horizontal load problem on a pile, the PMT is favored to obtain the soil data. The PMT in this case is a mini PMT called the Pencel (Fig. 30) which is driven in place or driven in a predrilled slightly smaller diameter hole if the soil is hard. As a result of many static and impact horizontal load tests at various scales (Lim, 2011, Mirdamadi, 2013), the following recommendations are made for the input parameters. EI

M s  0.036 B

C  N.s / m K  2.3Eo

2

PL g



 240 PL  kPa  and pu  pL

1

(60) 2 (61) (62)

Where B is the pile width, pL the PMT limit pressure, g the acceleration due to gravity, and Eo the first load PMT modulus. EQUIPMENT

3

TEST

4

Figure 30. Mini pressuremeter test

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parameters shown in Table 7. PMT tests were performed with a Pencel pressuremeter by first driving a slightly smaller diameter rod in the very stiff clay and then driving the Pencel probe in the slightly undersized hole. A comparison between the measured and calculated behavior of the pile (movement, load, and time) is presented in Fig. 32. The calculations were based on the simple Excel program (TAMU-POST, Mirdamadi, 2013) and a 4D FEM simulation using LS-DYNA (2006). The load was obtained by measuring the deceleration of the truck by placing an accelerometer on the bed of the truck and the movement by using high speed cameras. Table 7. PMT results by driven Pencel pressuremeter 5

DEPTH OF TEST

MODULUS

LIMIT PRESSURE

1m 1.8 m

45 MPa 25 MPa

1400 kPa 1200 kPa

a. STATIC TEST: LOAD VS. MOVEMENT DISPLACEMENT/WIDTH (/B) 0.12

0.16

LOAD (kN)

6

0.20 12.0

120

9.0

80

6.0

40

3.0

0 0

14

28

42

56

(kPa)

0.08

PRESSURE (P/BDv)

0.04

N k 7 x . 0 9 0 2 3 1 1 = x 8 5 3 . 7 0 . 0 x = u5 3 o. H 0

0.00 160

0.0 70

DISPLACEMENT (mm)

b. IMPACT TEST: MOVEMENT VS. TIME 1000

x DISPLACEMENT (mm)

800

7

600 Experiment TAMU-POST (Excel) LS DYNA

400 200 0 0.00

0.05

0.10

0.15

0.20

0.25

TIME (sec)

c. IMPACT TEST; FORCE VS.TIME 500

Experiment TAMU-POST (Excel) LS-DYNA

LOAD (kN)

400

8

300 200 100

Figure 31. Pick-up truck impact test Fig. 31 shows a photo sequence of an impact test where a 2300 kg pick up truck impacted a pile at 97.2 km/h. The pile was a steel pipe with a 356mm diameter and a 12.7mm wall thickness. It was embedded 2 m into a very stiff clay which gave the PMT

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0 0.00

0.05

0.10

0.15

TIME (sec)

0.20

0.25

0.30

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

b. PMT CHART BASED ON CORRELATION WITH CPT (adapted from Robertson and Wride, 1998)

d. IMPACT TEST: FORCE VS. MOVEMENT Static Experiment TAMU-POST (Excel) LS-DYNA

500

LOAD (kN)

400 300 200 100 0 0

200

400

600

800

1000

x DISPLACEMENT (mm)

Figure 32. Pick-up truck impact test results

Figure 33. Preliminary liquefaction charts based on the pressuremeter limit pressure

12 LIQUEFACTION CHARTS Liquefaction charts have been proposed over the years to predict when coarse grained soils will liquefy. In those charts (Fig. 33), the vertical axis is the cyclic stress ratio CSR defined as τav / σ’ov where τav is the average shear stress generated during the design earthquake and σ’ov is the vertical effective stress at the depth investigated and at the time of the in situ soil test. On the horizontal axis of the charts is the in situ test parameter normalized and corrected for the effective stress level in the soil at the time of the test. There is a chart based on the normalized SPT blow count N1-60 (Youd and Idriss, 1997). There is another chart based on the normalized CPT point resistance qc1 (Robertson and Wride, 1998). Using the correlations in Table 4, it is possible to transform the SPT and CPT axes into a normalized PMT limit pressure axis as shown in Fig. 34. The normalized limit pressure pL1 is

 p  pL1  pL  'a    ov 

0.5

(63)

Where pL is the PMT limit pressure, pa is the atmospheric pressure, and σ’ov is the vertical effective stress at the depth of the PMT test. Note that the data points on the original charts are not shown on the PMT chart not to give the impression that measurements have been made to prove the correctness of the chart. Some degree of confidence can be derived from the fact that the two charts give reasonably close boundary lines. Nevertheless, these two charts are very preliminary in nature and must be verified by case histories.

13 ANALOGY BETWEEN PMT CURVE AND EARTH PRESSURE-DEFLECTION CURVE FOR RETAINING WALLS The load settlement curve method for shallow foundations shows how one can use the PMT curve to predict the load settlement curve of a shallow foundation. This load settlement curve method was extended to the case of horizontally loaded piles. Can a similar idea be extended to the earth pressure versus deflection curve for retaining walls? One of the issues is that the PMT is a passive pressure type of loading so the potential for retaining walls may be stronger on the passive side than on the active side. Another issue is that the PMT test is a cylindrical expansion while the retaining wall is a plane strain problem. Fig. 34 shows the curves generated by Briaud and Kim (1998) based on several anchored wall case histories. The earth pressure coefficient K was obtained as the mean pressure p on the wall divided by the total vertical stress at the bottom of the wall. The mean pressure p was calculated by dividing the sum of the lock-off loads of the anchors by the tributary area of wall retained by the anchors. For each case history the lock off loads were known and the deflection of the wall was measured. Then the data was plotted with K on the vertical axis and the horizontal deflection at the top of the wall divided by the wall height on the horizontal axis. The shape of the curve is very similar to the shape of a PMT curve and a transformation function like the Γ function for the shallow foundation may exist but this work has not been done.

a. PMT CHART BASED ON CORRELATION WITH SPT (adapted from Youd and Idriss, 1997)

Figure 34. Earth pressure coefficient vs. wall deflection (after Briaud, Kim, 1998).

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14 CONCLUSIONS The purpose of this contribution was to show how the use of the PMT can be expanded further than current practice. In a first part, it is shown that more soil testing should take place in geotechnical engineering to reach a reasonable target of reliability. Then, it is theoretically demonstrated that if the lack of tensile resistance of soils is taken into account, the true soil modulus in compression is higher than what is obtained from conventional PMT data reduction. Then a procedure is investigated to recreate by hyperbolic extension the small strain early part of the curve lost by the decompression-recompression process associated with the preparation of the PMT borehole. The limitations of that procedure are identified. Best practice for preparing the PMT borehole, commonly expected values of PMT parameters, and correlations with other soil parameters are given. Reasoning is presented against the general use of the PMT unload reload modulus. It is shown that instead of limiting the use of the PMT test results to the modulus and the limit pressure, the entire expansion curve can be used to predict the load settlement behavior of shallow foundations and the load displacement curve of deep foundations under horizontal loading. Long term creep loading and cyclic loading are addressed. A solution is presented for the design of piles subjected to dynamic vehicle impact. It is also shown how the PMT can be very useful for the foundation design of very tall structures. Finally an attempt is made to generate preliminary soil liquefaction curves base on the normalized PMT limit pressure. 15 ACKNOWLEDGEMENTS The author wishes to thank the following individual for contributing to this paper: Roger Failmezger and Art Stephens for sharing some PMT data in sand, Ken Tand for sharing some plate load test data in stiff clay, Harry Poulos for providing some information on the Burj Khalifa measurements, Chris Haberfield for providing some information on the Nakheel Tower design, Clyde Baker for providing some information on his experience with the PMT and highrise foundation design. Several of my PhD students at Texas A&M University also contributed to this paper by making computations, preparing figures, formatting the manuscript, and more importantly discussing various aspects of the new contributions in this paper. They are: Alireza Mirdamadi, Ghassan Akrouch, Inwoo Jung, Seokhyung Lee. 16 REFERENCES 1. Baguelin F., Jezequel J.-F., Shields D.H., 1978, “The Pressuremeter and Foundation Engineering”, Trans Tech Publications, Clausthal-Zellerfeld, W. Germany, 1978. 2. Baker C.N. Jr., 2010, “Uncertain Geotechnical Truth and Cost Effective High-Rise Foundation Design”, 2009 Terzaghi Lecture, in Art of Foundation Engineering Practice, Edited by Mohamad H. Hussein; J. Brian Anderson; William M. Camp, Geotechnical Special Publications (GSP) 198, ASCE, Washington, USA. 3. Baker C.N., 2005, “The use of the Menard pressuremeter in innovative foundation design from Chicago to Kuala Lumpur”, the 2nd Menard Lecture, Proceedings of the 5th Int. Symp. on the Pressuremeter – ISP5, Paris, France, Presses de l’ENPC. 4. Baud J.-P., Gambin M., Schlosser F., 2013, “Courbes hyperboliques contrainte–déformation au pressiomètre Ménard autoforé”, Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013, Presses des Ponts et Chaussees, Paris, France. 5. Briaud J.-L., 1985, “ Pressuremeter tests at Amoco refinery”, consulting report to K.E. Tand and Associates, Houston, Texas.

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6. Briaud J.-L., 1992, “The Pressuremeter”, Taylor and Francis, London, pp.422. 7. Briaud J.-L., 1997, “SALLOP: Simple Approach for Lateral Loads on Piles”, Journal of Geotechnical and Geoenvironmental Engineering, Vol. 123, No.10, ASCE, Washington, USA. 8. Briaud J.-L., 2007, “Spread Footings in Sand: Load Settlement Curve Approach”, Journal of Geotechnical and Geoenvironmental Engineering, Vol 133, Issue 8, August 2007, ASCE, Reston, Virginia, USA. 9. Briaud J.-L., 2013, “Geotechnical Engineering: unsaturated and saturated soils”, John Wiley and Sons, New York, pp.848. 10. Briaud J.-L., Gibbens R., 1999, “Behavior of Five Spread Footings in Sand,” Journal of Geotechnical and Geoenvironmental Engineering, Vol. 125, No.9, pp. 787797, September 1999, ASCE, Reston, Virginia. 11. Briaud J.-L., Kim N.K., 1998, “Beam Column Method for Tieback Walls”, Journal of Geotechnical and Geoenvironmental Engineering, Vol. 124, No. 1, ASCE, Washington, DC. 12. Briaud J.-L., Makarim C.A., Little R., Tucker L., 1985, “Development of a pressuremeter method for predicting the behavior of single piles in clay subjected to cyclic lateral loading”, Research Report RF5112 to Marathon Oil Company, McClelland Engineers, Raymond International Builders, Shell Development Company, Dpt of Civil Engineering, Texas A&M University, College Station, Texas, USA, pp214. 13. Brown D.A., Reese L.C., 1985, “Behavior of a large scale pile group subjected to cyclic lateral loading”, Report to MMS, FHWA, and USAE-WES, Geotechnical Engineering Center Report GR85-12, Bureau of Engineering Research, Austin, Texas, USA. 14. Cox W.R., Dixon D.A., Murphy B.S., 1983, “Lateral load tests on 25.4 mm diameter piles in very soft clay in side by side and in line groups”, ASTM Special Technical Publication no. STP 835, pp 122-140. 15. Duncan, J.M., and Chang, C.Y. (1970) “Non-linear analysis of stress and strain in soils,” J. Soil Mech. Founds Div., ASCE, 96(SM5) 1629-1653. 16. Fascicule 62, 1993, “Regles techniques de conception et de calcul des foundations des ouvrages de genie civil”, Ministere de l’equipement, du logement, et des transports, Publications Eyrolles, Paris, pp182. 17. Haberfield C.M., Paul D.R., 2010, “Footing design of the Nakheel Tower, Dubai, UAE”, Proceedings of the Deep Foundation Conference, February 2010, Dubai, UAE, Deep Foundation Institute, 18pp. 18. Hossain, K. M. 1996. “Load settlement curve method for footings in sand at various depths, under eccentric or inclined loads, and near slopes.” Ph.D. thesis, Texas A&M Univ., Dept. of Civil Engineering, College Station, Tx, USA. 19. Jeanjean, P. 1995. “Load settlement curve method for spread footings on sand from the pressuremeter test.” Ph.D. dissertation, Texas A&M Univ., Dept. of Civil Engineering, College Station, Tx, USA. 20. Lim S.G., 2011, “Development of design guidelines for soil embedded post systems using wide flange I-beams to contain truck impact”, PhD dissertation, Zachry Dpt. of Civil Engineering, Texas A&M University, College Station, Texas, USA, pp394. 21. Little R.L., Briaud J.-L., 1988, “Full scale cyclic lateral load tests on six piles in sand”, Miscellaneous paper GL-88-27, US Army Engineer Waterways Experiment Station (now ERDC), Vicksburg, MS, USA, pp175. 22. LS-DYNA, 2006, “Theory Manual and User’s Manual version 971”, Livermore Software Technology Corporation, Livermore, CA. 23. Mayne, P. W., and G. J. Rix, 1993, “Gmax – qc Relationships for Clays,” Geotechnical Testing Journal, ASTM, Vol. 16, No. 1, pp. 54-60. 24. Mirdamadi A., 2013, “Deterministic and probabilistic model of single pile under lateral impact”, PhD dissertation, Zachry Dpt. of civil engineering, texas A&M university, College Station, Texas, USA.

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

32. Robertson, P.K., and Wride, C.E., (1998), ―Evaluating Cyclic Liquefaction Potential using the Cone Penetration Test,� Canadian Geotechnical Journal, Vol. 35, pp. 442459. 33. Samtini N.C., Liu J.-L., 2005, “Use of in situ tests to design drilled shafts in dense and cemented soils”, Proceedings of the Geo-Institute GeoFrontiers Conference, Austin, Texas, as part of the Mike O’Neill Memorial Volume, ASCE, Washington, DC, USA, 15pp. 34. Seed .B., Wong R.T., Idriss I.M., Tokimatsu K., 1986, “Moduli and damping factors for dynamic analyses of cohesionless soils, Journal of Geotechnical Engineering, ASCE, Vol. 112, GT11, pp1016-1032. 35. Tand K.E., 2013, “Plate load test results at Amoco refinery”, Personal communication. 36. Youd, T.L. and Idriss, I.M., (1997). ―Proceedings of the NCEER Workshop on Evaluation of Liquefaction Resistance of Soils�, Salt Lake City,UT, January 5-6, 1996, Technical Report NCEER-97-0022, National Center for Earthquake Engineering Research, University at Buffalo.

25. Morrison C., Reese L.C., 1986, “A lateral load test of a full scale pile group in sand”, Report to MMS, FHWA, and USAE-WES, Geotechnical Engineering Center Report GR85-12, Bureau of Engineering Research, Austin, Texas, USA. 26. NF P94-262, 2012, “Norme francaise, Justification des ouvrages geotechnicques, norme d’application nationale de l’Eurocode 7, foundations profondes, ISSN 0335-3931, AFNOR, pp206. 27. O’Neill M.W., 1983, “ “Group action in offshore piles”, ASCE Specialty Conference on Geotechnical Engineering in Offshore Engineering, Austin, Texas, USA. 28. O’Neill, M. W., Sheikh, S. A., 1985, “Geotechnical Behavior of Underreams in Pleistocene Clay,” Drilled Piers and Caissons II, ed. by C. N. Baker, Jr., ASCE, May, pp 57 – 75. 29. Poeppel A. R., 2013, Personal Communication, April 2013, Langan Engineering. 30. Poulos H.G., 2009, Tall buildings and deep foundations – Middle East challenges”, Terzaghi Oration, Proceedings of the 17th International Conference on Soil Mechanics and Geotechnical Engineering, Alexandria, Egypt, IOS Press publisher, 3173-3205 pp. 31. Rix, G.J. and Stokoe, K.H. (1991). Correlation of initial tangent modulus and cone resistance, Int Symp on Calibration Chamber Testing, Elsevier, New York, pp 351362.

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Rowe Rowe Lecture lecture The role of diffusion in environmental geotechnics Conférence Rowe Le rôle de la diffusion en géotechnique environnementale Shackelford C. Colorado State University, Fort Collins, Colorado, USA ABSTRACT: Diffusion of contaminants can play a significant if not dominant role in many applications encountered within the field of environmental geotechnics. The objective of this paper is to provide an overview of the important role diffusion plays in such applications. The presentation proceeds from a historical perspective, beginning with the recognition in the late 1970s to early 1980s that diffusion may be an important process in assessing contaminant migration through low permeability barriers in waste containment applications. Data from the literature and simplified model simulations are used to illustrate under what conditions diffusion is important, and the significance of diffusion with respect to different barrier components and types of barriers in waste containment applications is illustrated. The barriers considered include natural clays, compacted clay liners, geomembrane liners, geosynthetic clay liners, composite liners, vertical cutoff walls, subaqueous caps for contaminated sediments, and highly compacted bentonite buffers for high level radioactive waste containment. The significance of semipermeable membrane behavior on liquid-phase diffusion through bentonite-based barriers also is highlighted. The potential importance of matrix diffusion as an attenuation mechanism for contaminant transport also is illustrated, and the roles of both liquid-phase and gas-phase diffusion under unsaturated conditions are discussed. Finally, the role of diffusion in terms of remediation applications is illustrated via an example analysis illustrating the impact of reverse matrix or back diffusion on the effectiveness of pump-and-treat remediation. RÉSUMÉ: La diffusion de contaminants peut jouer un rôle significatif si ce n’est dominant dans le domaine de la géotechnique environnementale. L’objectif de cet article est de fournir une vue d’ensemble du rôle important de la diffusion dans de telles applications. La présentation suit une perspective historique : elle commence avec la reconnaissance vers la fin des années 70 au début des années 80, du fait que la diffusion peut être un processus important dans l’évaluation de la migration de contaminants à travers des barrières à perméabilité réduite dans des applications de confinement de déchets. Des données tirées de la littérature et des simulations avec des modèles simplifiés sont utilisées pour mettre en lumière sous quelles conditions la diffusion est importante. L’importance de la diffusion pour divers matériaux de barrières et types de barrières dans les applications de confinement des déchets est illustrée ; les barrières considérées comprennent les argiles naturelles, les liners d’argile compactée, les liners en géomembrane, les liners d’argile géosynthétique, les liners composites, les murs de confinement verticaux, les couvertures subaquatiques pour sédiments contaminés, et des zones tampons en bentonite fortement compactée pour le confinement des déchets radioactifs. L’importance du comportement des membranes semi-perméables sur la diffusion en phase liquide à travers des barrières à base de bentonite, telles que les liners en argile géosynthétique, est aussi présentée. L’importance potentielle de la diffusion en matrice en tant que mécanisme d’atténuation pour le transport de contaminants est aussi illustrée, et les rôles de la phase liquide comme de la phase gazeuse dans des conditions non saturées sont examinés. Finalement, le rôle de la diffusion en terme d’applications de dépollution est illustré via l’analyse d’un exemple qui décrit l’impact de la diffusion arrière sur l’efficacité de la dépollution « pump-and-treat » (pompage-écrémage-filtration). KEYWORDS: Advection; Containment; Contaminant; Diffusion; Fick's laws; Membrane behavior; Remediation 1 INTRODUCTION The advent of the formal sub-disciplinary field of geotechnical engineering known as environmental geotechnics can be traced to the early to mid 1970s, soon after the formation of environmental regulatory agencies, such as the United States Environmental Protection Agency (US EPA) formed in 1970, whose purpose was to enforce environmental regulations promulgated for the protection of human health and the environment (Shackelford 1999, 2000). One of the first orders of business for these regulatory agencies was to provide guidelines and regulations for the safe disposal of a variety of liquid and solid wastes, including hazardous solid waste (HSW) and municipal solid waste (MSW). For example, the Resource Conservation and Recovery Act (RCRA) promulgated in the US in 1976 provided detailed guidelines for the use of low permeability barriers of recompacted clay, since known as compacted clay liners

(CCLs), to minimize the migration of liquids and contaminants emanating from HSW and MSW in the form of RCRA Subtitles C and D, respectively. Until this period of time, wastes had been disposed largely with relatively little or no regard for any potential environmental consequences, often in unlined pits and dumps or in facilities that relied primarily upon the inherent low permeability of any natural soil within the vicinity of the disposal location. Because of the lack of concern for environmental consequences resulting from waste disposal prior to this period, contamination at numerous disposal sites (hundreds to thousands) had already occurred over the previous decades, such as the infamous Love Canal site located in Niagara Falls, New York, USA. Public awareness of the potential environmental health concerns from such existing contamination resulted in the realization of the need to clean up or remediate the existing contamination from sites that had already been polluted. An example of this

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where Jd is the diffusive mass flux, or the rate of change in mass of the chemical species per unit cross sectional area perpendicular to the direction of diffusion [ML-2T-1; M = units of mass, L = units of length, and T = units of time], n is the total porosity of the porous medium, D* is the effective diffusion coefficient [L2T-1], a (< 1) is the apparent tortuosity factor [-], Do is the aqueous-phase or free solution (without porous medium) diffusion coefficient [L2T-1], and ic is the concentration gradient in the direction of diffusion [-], which is positive when directed towards decreasing solute concentration. The apparent tortuosity factor, a, represents the product of the actual matrix tortuosity factor representing the geometry of the interconnected pores, m (< 1), and the restrictive tortuosity factor, r, as follows (Malusis and Shackelford 2002a, Shackelford and Moore 2013):

realization is the promulgation in the US in 1980 of the law known as the Comprehensive Environmental Response, Compensation, and Liability Act (CERCLA), also known as Superfund, that authorized the US EPA to respond to releases, or threatened releases, of hazardous substances that may endanger public health, welfare, or the environment, and also enabled the US EPA to force parties responsible for environmental contamination to clean up such contamination and/or to reimburse the Superfund for response or remediation costs incurred by the government. Thus, the burgeoning field of environmental geotechnics began to address technical issues related not only to the prevention of contamination resulting from disposal of new waste, but also to the remediation or clean up of existing contamination resulting from improper disposal practices in the past. Because of the experience of geotechnical engineers in using compacted clays for applications such as the low permeability cores of engineered earthen dams (e.g., Mitchell et al. 1965), geotechnical engineers immediately became involved and identified with the design and use of CCLs as engineered barriers for disposal of new wastes. However, the early emphasis in the use of CCLs as barriers for waste containment focused primarily on the physical and mechanical properties of the CCLs, such as minimizing the hydraulic conductivity, kh, of the CCL in order to reduce the rate of seepage of contaminated liquids (e.g., leachates), v, through the CCLs resulting from the application of a hydraulic gradient, ih, in accordance with Darcy's law (i.e., v = kh·ih). The realization of the need to consider the chemical properties of the contaminants as well as the potential detrimental impacts resulting from the physico-chemical interactions between the liquids being contained and the soils used to contain the liquids was more gradual, and has developed over an extended time frame. In particular, beginning in the late 1970s to early 1980s, diffusion became recognized as a potentially important process in assessing contaminant migration through low permeability barriers in waste containment applications. This recognition led to a progressively greater understanding of the role diffusion plays in a wide variety of applications in environmental geotechnics, including applications in both waste containment and remediation. Thus, the objective of this paper is to provide an overview of the role diffusion plays in the field of environmental geotechnics.

a mr

where r represents the product of all other factors that may be effective in reducing the diffusive mass flux of a chemical species, such as ion exclusion. In essence, r represents the ratio of the effective to total porosities, or (Shackelford and Moore 2013): n r  e n

(3)

where ne ≤ n such that r ≤ 1.The recognition of an effective porosity takes into account the possibility that that there may be pores that are not interconnected or are inaccessible to specific solutes such that only a fraction of the pore space may be available for diffusion (Shackelford and Moore 2013). Fick's second law governing transient one-dimensional diffusion of chemical species subject to first-order linear decay in porous media can be written as follows (e.g., Shackelford and Daniel 1991a, Shackelford and Rowe 1998, Shackelford and Moore 2013):

C D*  2C  2C    C Da  C t Rd x 2 x 2

(4)

where C is solute concentration [ML-3], Rd is the dimensionless retardation factor, Da (=D*/Rd) is the apparent diffusion coefficient [L2T-1], and  is the decay constant [T-1]. For chemical species subjected to first-order decay (e.g., radionuclides),  is inversely related to the half life of the chemical species, t1/2, such that  decreases as t1/2 increases. For this reason, the decay term in Eq. 4 can be (and often is) ignored without any significant loss in accuracy for chemical species with half lives that are considerably longer than the time frame being considered for diffusion (Shackelford and Moore 2013). The retardation factor in Eq. 4 accounts for linear, reversible, and instantaneous sorption of a chemical species, and represents the ratio of the total mass of chemical species per unit total volume of porous medium relative to the aqueous-phase mass of chemical species per unit total volume of porous medium. For water saturated porous media, Rd may be expressed as follows:

2 WHAT IS DIFFUSION? Diffusion is a fundamental, irreversible process whereby random molecular motions result in the net transport of a chemical species (e.g., ion, molecule, compound, radionuclide, etc.) from a region of higher chemical potential to a region of lower chemical potential (Quigley et al. 1987, Shackelford and Daniel 1991a, Shackelford and Moore 2013). Since chemical potential is directly related to chemical concentration, diffusion is more commonly described as the net transport of a chemical species due to a gradient in the concentration of the chemical species. The mass flux of a chemical species in a porous medium due to diffusion can be described by Fick's first law, which for one-dimensional diffusion may be written as follows (e.g. Shackelford and Daniel 1991a, Shackelford and Rowe 1998): J d nD*ic n  a Do  ic

(2)

 Rd  1  d K d n

(1)

2

128

(5)

Honour Lectures / Conférences honorifiques

where d is the dry density of the solid phase, or mass of solids per unit total volume of solids [ML-3], and Kd is the distribution coefficient [L3M-1], which relates the solidphase concentration, Cs, expressed as the sorbed mass of the chemical species per unit mass of the solid phase [MM1 ], to the aqueous-phase concentration, C, of the chemical species (i.e., assuming linear, reversible, and instantaneous sorption), or Kd = Cs/C. As a result, for sorbing chemical species, Kd > 0 such that Rd > 1, whereas for nonsorbing chemical species, Kd = 0 (i.e., Cs = 0) such that Rd = 1. Thus, Da as given by Eq. 4 represents a lumped effective diffusion coefficient that includes the effect of attenuation via Rd. For this reason, Da also has been referred to as the effective diffusion coefficient of a reactive chemical species (Shackelford and Daniel 1991a). For water unsaturated porous media, the total porosity, n, in Eq. 5 is replaced by the volumetric water content, w, where w = nSw and Sw is the degree of water saturation (0 ≤ Sw ≤ 1). Since the notation for the various diffusion coefficients defined herein may not match the notation used by others (e.g., D* as defined herein also is commonly designated as De), caution should be exercised in terms of understanding the basis for the definition of the various diffusion coefficients when interpreting values extracted from the published literature. Unless indicated otherwise, the default definition of the diffusion coefficient used herein is that corresponding to D*. For liquid-phase diffusion of aqueous soluble chemical species in saturated porous media, values of D* generally fall within range 10-9 m2/s > D* > 10-11 m2/s, with lower values of D* being associated with finer textured and/or denser soils (Shackelford and Daniel 1991a, Shackelford 1991). Since a < 1, the upper limit on D* of 10-9 m2/s is dictated by the Do values, which generally ranges from about 1 to 2 x 10-9 m2/s for most aqueous soluble chemical species, except for those involving H+ or OH-, in which case Do is approximately 2 to 4 times higher (Shackelford and Daniel 1991a). Values of D* < 10-11 m2/s are possible in situations involving bentonite-based containment barriers, such as highly compacted bentonite buffers for high-level radioactive waste disposal, primarily as a result of ion exclusion resulting from the existence of semipermeable membrane behavior such that r < 1 (e.g., Malusis and Shackelford 2002a, Shackelford and Moore 2013). Liquid-phase values of D* for unsaturated porous media generally decrease with decreasing w or Sw and can be several orders of magnitude lower than the respective values at full water saturation (Shackelford 1991). Finally, values of Da for reactive chemical species (e.g., heavy metal cations) typically range from one to several orders of magnitude lower than the corresponding D* values due to attenuation mechanisms (e.g., sorption, ion exchange, precipitation, etc.), i.e., Rd > 1. 3 WHEN IS DIFFUSION SIGNIFICANT? Following the approach of Shackelford (1988), the significance of diffusion on the migration of aqueous soluble chemical species, or solutes, through porous media can be illustrated with the aid of solute breakthrough curves, or BTCs, representing the temporal variation in the concentration of a given chemical species at the effluent end of a column of porous medium. As depicted schematically in Fig. 1a, BTCs can be measured in the laboratory for a column of a porous medium of length L by (a) establishing steady-state seepage conditions, (b) continuously introducing at the influent end of the column

a chemical solution containing a known chemical species at a concentration Co, and (c) monitoring the concentration of the same chemical species emanating from the column as a function of time, or C(L,t) (Shackelford 1993, 1994, 1995, Shackelford and Redmond 1995). Because the source concentration, Co, is constant, the BTCs typically are presented in the form of dimensionless relative concentration, C(L,t)/Co, versus elapsed time. The time required for the solute to migrate from the influent end to the effluent end of the column is referred to as the "breakthrough time" or the "transit time." For example, consider the three BTCs depicted in Fig. 1b for the case of a low permeability clay (kh = 5 x 10-10 m/s) contained within a column of length 0.91 m and at a porosity of 0.5, and subjected to an applied hydraulic gradient, ih, of 1.33. The chemical solution serving as the permeant liquid contains a nonreactive solute at a constant concentration of Co and is assumed to be sufficiently dilute such that no adverse interactions between the clay and the solution result in any changes in kh during the test. The BTC in Fig. 1b labeled "pure advection" represents the case commonly referred to as "piston" or "plug" flow, whereby the breakthrough time is the time predicted in the absence of any dispersive spreading of the solute front using the seepage velocity, vs, in accordance with Darcy's law (i.e., t = L/vs = nL/khih). Under purely advective (hydraulic) transport conditions, 21.8 yr would be required for the solute to completely break through the effluent end of the column (i.e., C(L,t)/Co = 1) in the absence of any dispersive spreading of the solute front, owing to the very low seepage rate. The BTC in Fig. 1b labeled "advection plus mechanical dispersion" represents the spreading effect on the solute front primarily due to mechanical (advective) dispersion (i.e., diffusive dispersion is assumed negligible), which is the case commonly depicted in groundwater hydrology textbooks because the primary concern pertains to contaminant migration within aquifers, or coarse-grained, water-bearing strata subjected to relatively high seepage velocities. The BTC for this case, as well as that for the next case, was generated using a commonly applied analytical model to the advective-dispersive solute transport equation developed by Ogata and Banks (1961) for the stated conditions of the column test (e.g., Shackelford 1990). In this case, the dispersive spreading of the solute front is attributed to variations in the porescale velocity profiles at the column scale and heterogeneities in hydraulic conductivity at the field scale (e.g., Shackelford 1993). Due to this spreading effect of the solute front, there are an infinite number of possible breakthrough times depending on the value of C(L,t)/Co used to define the breakthrough time. However, the typical practice is to evaluate the breakthrough time at a relative concentration of 0.5, which is the time at which the BTCs for pure advection and advection plus mechanical dispersion intersect. The BTC in Fig. 1b labeled "advection plus diffusion" is the true BTC for this column, as this BTC reflects the situation when the seepage velocity is sufficiently low such that the effect of diffusion is not masked by the effects of advection and mechanical dispersion. The spreading effect is still noticeable in this BTC, but this BTC is displaced to the left of the previous two BTCs, resulting in a breakthrough time at C(L,t)/Co of 0.5 of 14.8 yr, which is considerably less than the value of 21.8 yr for the two previous cases where diffusion is ignored. Thus, failure to include the diffusion as a transport process under the 3 129

Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

conditions of this column test would result in not only an incorrect but also an unconservative (high) estimate of the breakthrough time.

1

0.8 0.7

Pure Advection

(b)

Advection + Diffusion

0.6

Advection + Mechanical Dispersion

0.5 21.8 yr

0.4 14.8 yr

Relative Concentration, C(L,t)/C

o

0.9

0.3 0.2 0.1

-10

k = 5 x 10 h

m/s

L = 0.91 m n = 0.5 i = 1.33 h

0 0

10

20 30 40 50 Elapsed Time, t (yr)

60

-8

(c)

L = 0.91 m n = 0.5 i = 1.33

-10

10

-11

Pure Diffusion (i = 0)

Pure Advection

h

48.5 yr

-9

10

14.8 yr 21.8 yr

h

h

Hydraulic Conductivity, k (m/s)

10

k = 2.2 x h

Advection + Diffusion

10

-10

m/s

10

0 10 20 30 40 50 60 70 80 Transit Time @ C(L,t)/C = 0.5, t (yr) o

0.5

Figure 1. Effect of diffusion on solute transport through a column of soil of length L, porosity n, and hydraulic conductivity kh, under a hydraulic gradient of ih; (a) column containing porous medium; (b) breakthrough curves illustrating effect of diffusion at low kh; (c) transit (breakthrough) times, t0.5, as a function of kh (modified after Shackelford 1988).

As previously implied, the decrease in the breakthrough time due to diffusion evident in the BTCs shown in Fig. 1b is a function of the magnitude of the seepage velocity. This dependence on vs is illustrated in Fig. 1c, where the breakthrough times at C(L,t)/Co of 0.5, or t0.5, are shown for the cases of pure advection and advection plus diffusion as a function of the kh of the porous medium in the column, all other conditions being the same (i.e., L = 0.91 m, n = 0.5, ih = 1.33). The limiting case of pure diffusion (ih = 0) also is shown in Fig. 1c for comparison. The horizontal distance between the pure advection and advection plus diffusion curves represents the offset distance at C(L,t)/Co of 0.5, or t0.5, in Fig. 1b for a given kh. The independence of pure diffusion on kh is represented by a vertical line corresponding to t0.5 of 48.5 yr. Three observations are apparent from the curves shown in Fig. 1c (Shackelford 1988): (1) diffusion has an effect (i.e., t0.5 > 0) even at a kh of 10-9 m/s, which typically is the maximum

regulated kh value for many waste containment applications; (2) the sole use of Darcy's law (i.e., pure advection) to predict breakthrough times is extremely unconservative at kh values less than about 2.2 x 10-10 m/s; and (3) diffusion starts to become the dominant transport process (i.e., as the curve for advection plus diffusion starts to approach asymptotically that for pure diffusion) at a kh value of about 2 to 3 x 10-10 m/s. Of course, the solute mass flux also would be significantly reduced with decreasing kh, but still may be environmentally significant (e.g., Johnson et al. 1989). Regardless, this simplified analysis illustrates the importance of diffusion in low permeability porous media. In terms of concentration profiles, consider the scenario depicted in Fig. 2a corresponding to a ponded source of liquid containing a nonreactive chemical species at a constant concentration, Co, underlain by an initially uncontaminated soil with an n of 0.5. The resulting concentration profiles beneath the source at an elapsed time of 5 yr assuming a D* of 6 x 10-10 m2/s are shown in Figs. 2b, 2c, and 2d for advective dominated (kh = 10-8 m/s), diffusive significant (kh = 10-9 m/s), and diffusive dominated (kh = 10-10 m/s) cases, respectively. For the advective dominated case (Fig. 2b), the pure advective (seepage) front extends the furthest distance (> 4 m), there is little difference between mechanical dispersion and diffusion, and all concentration profiles intersect at C(x,t)/Co of 0.5. For the diffusive significant case (Fig. 2c), the pure advective front is much shallower (< 1 m), dispersion due to diffusion is much greater than that due to mechanical dispersion, and the concentration profile for advection plus diffusion intersects that for pure advection at C(x,t)/Co ~ 0.68. Finally, for the diffusive dominant case (Fig. 2d), the depth of penetration of the pure advective front is virtually imperceptible as is the concentration profile for advection plus mechanical dispersion, and almost the entire concentration profile for advection plus diffusion extends beyond that for the pure advection case. Thus, although the extent of contaminant migration is greatest when the kh value of the subsurface soil is the greatest, the extent of migration predicted on the sole basis of advection (i.e., Darcy's law) becomes increasingly unconservative as the kh of the subsurface soil decreases, such that diffusion becomes more prominent. As will be shown subsequently, associating the shapes of concentration profiles with the dominant transport processes played an important role in the recognition of diffusion as a potentially important transport process. 4 DIFFUSION IN CONTAINMENT APPLICATIONS 4.1 Containment Scenarios In terms of waste containment scenarios, there are three general scenarios of interest, as illustrated in Fig. 3. The first case illustrated in Fig. 3a is the limiting case of pure diffusion. For waste containment scenarios involving horizontal barriers (liners), the likelihood that the pure diffusion case will be realized in practice is relatively remote, as there almost always will be a hydraulic gradient driving advective transport.

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Honour Lectures / Conférences honorifiques

Depth (m)

R e la tiv e C o n c e n tra tio n , C (x ,t)/C o 0 0 .2 0 .4 0 .6 0 .8 1 0 A d v e c tio n + 1 A d v e c tio n D iffu s io n + 2 3

M e c h a n ic a l D is p e rs io n P u re A d v e c tio n

4 5 k 6

h

= 10

-8

m /s

(b )

R e la t iv e C o n c e n tr a t io n , C (x , t )/ C 0 0 . 2 0 . 4 0 .6 0 .8 1 o 0

Depth (m)

1 2 3 4

P u re A d v e c t io n A d v e c t io n + M e c h a n ic a l D is p e rs io n A d v e c t io n + D if f u s io n

Figure 3. Transport scenarios across horizontal barriers for waste containment: (a) pure diffusion; (b) diffusion with positive (outward) advection; (c) diffusion with negative (inward) advection (modified after Shackelford 1989, 1993).

The third scenario (Fig. 3c) pertains to the case where the hydraulic and concentration gradients act in opposite directions, such that advective transport is directed inward towards the containment side of the barrier, whereas diffusive transport still is directed outward. As a result, the net outward advance of the chemicals is slowed or "retarded" by the opposing hydraulically driven transport. This situation would arise, for example, when the containment system is located at a site with a high groundwater table, such as a perched water table, such that the barrier is located below the water table. This scenario has been referred to as "zone of saturation" containment (e.g. Shackelford 1989, 1993). The scenario also has been referred to as a "hydraulic trap," because the inward directed hydraulic gradient enhances the containment function (e.g., Rowe et al. 2000, Badv and Abdolalizadeh 2004). However, because diffusion is still prevalent, the existence of an opposing hydraulic gradient does not necessarily mean that no contaminant will escape containment, as the net effect will depend on the magnitude of advective transport relative to that for diffusive transport. Also, the effectiveness of inward gradient landfills may not be as complete as expected in the case where the barrier possesses semipermeable membrane properties (Whitworth and Ghazifard 2009).

5 k 6

h

= 10

-9

m /s

(c )

R e l a t iv e C o n c e n t r a t io n , C ( x , t ) / C 0 0 .2 0 .4 0 .6 0 .8 1 0 P u re A d v e c tio n

Depth (m)

1 2 3 4

o

A d v e c t io n + M e c h a n ic a l D is p e r s i o n A d v e c t io n + D if f u s io n

5 k 6

h

= 10

-10

m /s

(d )

Figure 2. Representative concentration profiles beneath a ponded source of liquid after an elapsed time of 5 yr: (a) schematic of scenario (n = 0.5, D* = 6 x 10-10 m2/s, ih = 1.33); (b) advective dominated case; (c) diffusive significant case; (d) diffusive dominated case.

The most common scenario is illustrated in Fig. 3b, where both hydraulic and concentration gradients act in the same direction to drive advective and diffusive chemical transport from the containment side of the barrier to the surrounding medium. This scenario also is the scenario depicted previously with respect to Figs. 1 and 2.

4.2 Diffusion through Barriers or Barrier Components 4.2.1 Diffusion in Natural Clays The recognition that diffusion may play an important role in governing contaminant migration gained momentum in the late 1970s with the publication of a case study by Goodall and Quigley (1977) describing the field concentration profiles that existed beneath two landfill sites near Sarnia, Ontario, Canada, viz. the Confederation Road landfill and the Blackwell Road landfill. The pore water obtained from Shelby tube samples collected beneath a landfill sited directly on top of intact glacial till, and the 5

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

concentrations of the primary cations, i.e., K+, Na+, Ca2+ and Mg2+ were plotted as a function of depth beneath the interface of the waste and the till, as shown in Fig. 4. The kh of the silty clay till in the vicinity of the landfills was measured using both laboratory and field methods over depths ranging from 1.5 to 27.4 m, and 14 of the 18 measured kh values were lower than 10-10 m/s. At the Confederation Road landfill site (Fig. 4a), the landfill was located directly upon unfissured, intact, gray clay below a desiccated crust, and piezometers indicated downward seepage gradients that enabled, together with measured values of kh and n, calculating a maximum advective front of 0.04 m over the 6-yr life of the landfill. However, as shown in Fig. 4a, the measured disperse cation front extended to a much greater distance of about 0.3 m, well beyond the advective front. The authors recognized that advective migration in accordance with Darcy's law could not be the primary transport process, and that the cation concentration profiles resembled those that would be predicted on the basis of diffusive dominated conditions (e.g., compare Fig. 4a with Fig. 2d). The same conclusions were drawn with respect to the cation concentration profiles beneath the Blackwell Road landfill, although the diffusive front had extended to an even greater depth (0.4-0.8 m), despite the existence of upward hydraulic gradients resulting from consolidation of the underlying till due to loading by the overlying MSW. In this case, the authors attributed the greater extent of cation migration to the existence of fissures in the clay, providing pathways for more rapid downward migration, followed by diffusion of the cations into the surrounding intact clay matrix resulting in the observed concentration profiles. This latter process is referred to as "matrix diffusion" and will be discussed in more detail later. A subsequent study reported by Crooks and Quigley (1984) involving additional field analyses and associated laboratory testing confirmed the earlier conclusions drawn by Goodall and Quigley (1984). Another field study reported by Johnson et al. (1989) involved obtaining vertical core samples from an impervious, unweathered, water-saturated clay deposit beneath a 5-yr-old hazardous waste landfill site in southwestern Ontario, Canada. Sections of the cores were analyzed for chloride and volatile organic compounds (VOCs). Waste-derived chloride was detected in the clay to a maximum depth of ~ 0.83 m below the bottom of the landfill, whereas the most mobile VOCs were found to a depth of only ~ 0.15 m. The authors concluded that the downward transport of these chemical species was the result of simple Fickian diffusion and, more importantly, that the results of this study had important implications for clay-lined waste disposal sites. Specifically, they noted that for engineered clay liners of typical thickness of ~ 1 m, simple diffusion could cause breakthrough of mobile contaminants in approximately 5 yr, and that the diffusive flux emanating from such liners could be large, at least from the perspective of protection of human health and the environment.

Cation Concentration (mg/L) 100 200 300

0

(a) 5.5 K

+

Bottom of Landfill + 2+ Na Ca2+ Mg

Depth (m)

0.3 m ~ Extent of Diffusive Front

6

Glacial Till

6.5 Background

0 2

Cation Concentration (mg/L) 100 200 300 400 500 600 700 Rubber Boot, Plastic Bags, Milk Cartons, Wire Fence, Newspaper Bottom of Landfill +

Depth (m)

K

Na

+

(b)

2+

Mg

2.5

0.4 m Ca2+

~ Extent of Diffusive Front 3

Consolidation Flow Gradient

Glacial Till

Cation Concentration (mg/L) 100 200 300 400 500

0

(c) Bottom of Landfill +

K

Depth (m)

2.5

Na

+

Ca

2+

2+

Mg

0.4 m ~ Extent of Diffusive Front

3

Consolidation Flow Gradient Glacial Till

3.5

Figure 4. Cation concentration profiles beneath two landfills at Sarnia, Ontario, Canada: (a) Confederation Road landfill; (b) Blackwell Road landfill, borehole 1; (c) Blackwell Road landfill, borehole 3 (modified from Goodall & Quigley 1977).

As a result of these and other studies involving natural clays (e.g., Barone et al. 1989, Barone et al. 1992, Myrand et al. 1992, Sawatsky et al. 1997, Donahue et al. 1999, Itakura et al. 2003, Mieszkowski 2003, Appelo et al. 2008, Jakob et al. 2009), diffusion became recognized as an important transport process in low permeability porous media. This recognition led to studies focused on evaluating the role that diffusion played in terms of governing contaminant migration through engineered clay barriers, such as CCLs. The results of several of these studies are described in the following section. 4.2.2 Diffusion through Engineered Clay Barriers Although several studies have focused directly on evaluating the role of diffusion in governing contaminant migration through engineered clay barriers, such as CCLs used for MSW, HSW, and low-level radioactive waste (LLRW) disposal as well as compacted bentonite buffers 6

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used in high-level radioactive waste (HLRW) disposal, the vast majority of these studies have been laboratory scale studies (e.g., Crooks and Quigley 1984, Gillham et al. 1984, Shackelford et al. 1989, Shackelford and Daniel 1991b, Rowe and Badv 1996a,b, Cotten et al. 1998, Roehl and Czurda 1998, Foged and Baumann 1999, Headley et al. 2001, Rossanne et al. 2003, Çamur and Yazicigil 2005, Frempong and Yanful 2008, Hong et al. 2009, Korf et al. 2011, De Soto et al. 2012). By comparison, relatively few field-scale studies of diffusion in compacted clay barriers have been reported, primarily because the extent of contaminant migration under diffusion dominated conditions would not be sufficient within the operational time-frame of most barriers to allow for such evaluation without violating the integrity of the barrier via core sampling. However, two exceptions to this restriction are the Keele Valley Landfill (KVL) located north of Toronto in Maple, Canada, which was operational between 1984 and 2002 (Rowe 2005), and a field-scale CCL that was specifically constructed as a field research study on the campus of the University of Illinois to evaluate contaminant transport through CCLs and was operational for 13 yr (1988-2001) (Cartwright and Krapac 1990, Toupiol et al. 2002, Willingham et al. 2004). Concentration profiles existing across the interface of sand overlying the clay liner at the KVL after 4.25 yr of operation are shown in Fig. 5. The profiles in Fig. 5a are for chloride, whereas those in Fig. 5b pertain to a group of VOCs known as the BTEX compounds (benzene, toluene, ethylbenezene, and xylene). The liner generally was 1.2 m in thickness, with a kh that was regulated to be 10-10 m/s or less (King et al. 1993). The sand overlying the clay liner was meant primarily to be a protection layer for the underlying clay liner, and the upper portion of the sand layer became clogged within the first four years such that the sand layer did not contribute to the hydraulic performance of the leachate collection system (Rowe 2005). This clogging resulted in a lack of flow through the sand layer, such that the sand layer actually served as part of the low-permeability barrier system whereby diffusion was the dominant transport process. For example, Rowe (2005) reported that, in the case of the chloride concentration profile (Fig. 5a), the assumption of purely diffusive transport using a D* value for chloride of 6 x 1010 m2/s resulted in a predicted profile that matched the measured profile well, and that the concentration profiles for the BTEX compounds, especially toluene, through both the sand and the clay resembled those for diffusion dominated conditions. A detailed description of the construction and installation of monitoring for the prototype CCL constructed as a research project at the University of Illinois can be found in Cartwright and Krapac (1990). The compacted liner was approximately 0.9-m thick and was constructed using Batestown Till compacted wet of optimum water content. The dimensions of the liner facility were 10 m x 17 m x 1 m, which included an instrumented and ponded test area of 7.3 m x 14.6 m x 0.9 m (Willingham et al. 2004). The entire facility was enclosed within a heated shelter to minimize weather effects and prevent infiltration from rainfall. As part of the monitoring system, large-ring infiltrometers (LRI), 1.5 m in diameter were installed on the surface of the liner and subsequently filled to a depth of 0.295 m with water tagged with tracers (tritium, (HTO) and bromide (Br-)). Approximately one year later, the water level was raised to 0.31 m and maintained at that level for about 8.5 yr, and

Depth from Sand-Clay Interface (m)

then the water level was allowed to decrease due to evaporation and infiltration, but never reached the liner surface before the study was terminated (Willingham et al. 2004). Chloride Concentration, C (mg/L) 0 1000 2000 3000 4000 -0.5 (a) Municipal Solid Waste -0.4 -0.3 Black Sand -0.2 Reduced Gray Sand -0.1 Light Brown Sand Interface 0 Brown 0.1 Clayey 0.2 Liner Measured Data 0.3 0.4 0.5 Computed Diffusion Profile 0.6 -10 2 (D* = 6.5 x 10 m /s, t = 4.25 yr) 0.7 0.8

Depth from Sand-Clay Interface (m)

0 -0.5 -0.4 -0.3 -0.2 -0.1 0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8

VOC Concentration, C (g/L) 200 400 600 800 1000 (b)

Municipal Solid Waste Black Sand Reduced Gray Sand

Xylene

Interface

Light Brown Sand

Toluene Ethylbenzene

Brown Clayey Liner

Benzene

Figure 5. Concentration profiles within the engineered barrier system at the Keele Valley Landfill, Maple, Ontario, Canada: (a) chloride concentration profiles (modified after King et al. 1993); (b) concentration profiles for VOCs (modified after Rowe 2005).

A cross-sectional schematic for the LRI set-up is illustrated in Fig. 6a, and concentrations profiles of Br- as a function of depth and radial distance, r, from the centerline of the LRI are shown in Figs. 6b,c. The profiles were fitted with an analytical three-dimensional transport model to the advective-dispersive-diffusive transport equation. As shown in Figs. 6b,c, reasonable fits to the measured data were obtained for D* values varying from 3.0 x 10-10 m2/s to 8.0 x 10-10 m2/s. The authors concluded that: (a) Brtransport through the field-scale liner was controlled by diffusion, (b) the vertical and horizontal diffusion coefficients were the same, and (c) CCLs can be constructed as diffusion controlled barriers that are capable of mitigating chemical transport from localized leaks or source zones.

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0

0

Bromide Concentration, C (mg/L) 5 10 15 20 25 30 35 r=0

0.1

GML is via molecular diffusion, and the only contaminants that can diffuse substantially through the GML are those that can partition into the polymer comprising the GML, which generally limits the contaminants to organic compounds, such as VOCs. For example, Rowe (2005) reported the results of a long-term diffusion test involving a 2-mm-thick HDPE geomembrane subjected to a difference in NaCl concentration of 2.2 g/L, where the measured concentration of chloride on the downgradient side of the geomembrane after about 12 yr of exposure was only 0.02 % of the source concentration, which was within the range of the analytical uncertainty of the chemical analysis. Rowe (2005) also cites the results of an independent study that indicated negligible diffusion of heavy metals (Cd2+, Cu2+, Mn2+, Ni2+, Pb2+, Zn2+) from a 0.5 M acid solution (pH = 1-2) through an HDPE over a 4yr period. In this regard, there have been numerous studies evaluating diffusion of a wide variety of organic chemicals through a wide variety of different polymer-based GMLs (Rowe et al. 1995, Park and Nibras 1996, Park et al. 1996a,b, Xiao et al. 1996, Sangam and Rowe 2001a, Joo et al. 2004, 2005, McWatters and Rowe 2010, Jones et al. 2011, Saheli et al. 2011, Touze-Foltz et al. 2011). A primary outcome from most of these studies is that geomembranes formed from a single polymer, such as high density polyethylene (HDPE), linear low-density polyethylene (LLDPE), very low-density polyethylene (VLDPE), and polyvinyl chloride (PVC), typically provide little resistance to diffusion of VOCs (e.g., Edil 2003). In this regard, the general process for diffusion of such organic chemicals through GMLs in response to an aqueous-phase concentration difference, -C = Co – Ce > 0, established across a GML is illustrated schematically in Fig. 7 (e.g., see Rowe 1998, Katsumi et al. 2001). First, the organic chemical partitions from the external aqueous solution into the geomembrane (adsorbs) at a concentration KgCo, where Kg is the chemical-geomembrane partitioning coefficient. Second, the chemical diffuses through the geomembrane in response to a concentration difference within the GML of -Cg = KgCo – KgCe > 0, where KgCe has been established on the basis of the external aqueousphase concentration, Ce. Finally, the chemical partitions from the geomembrane (desorbs) back into the lower bounding aqueous solution.

(b) -11

v = 4.0 x 10 s

m/s

0.2 Depth (m)

0.3 0.4 0.5 Measured-Source

0.6

Measured-Soil

0.7

-10

D* = 3.0 x 10

-10

D* = 5.5x 10

0.8

-10

D* = 8.0 x 10

0.9

0

0

2

m /s 2

m /s

Bromide Concentration, C (mg/L) 5 10 15 20 25 30 35 40 r = 0.60 m

0.1

2

m /s

-10

v = 4.0 x 10 s

(c)

m/s

0.2 Depth (m)

0.3 0.4 0.5 0.6 0.7 0.8

Measured-Source Measured-Soil (r = 0.53 m) Measured-Soil (r = 0.63 m) -10

D* = 3.0 x 10

-10

D* = 5.5x 10

-10

D* = 8.0 x 10

2

m /s 2

m /s 2

m /s

0.9

Figure 6. Bromide concentration profiles with a prototype compacted clay liner: (a) schematic cross section of large-ring infiltrometer; (b) and(c) concentration profiles at radii of 0 and 0.60 m, respectively, from the centerline of the LRI (modified after Willingham et al. 2004).

4.2.3 Diffusion through Geomembrane Liners Geomembrane liners (GMLs) are thin (typically 0.76 mm to 3.05 mm) polymer-based materials that are commonly used as barriers or components of barrier systems for containment applications. In such applications, the only way for aqueous-phase inorganic contaminants to migrate through the polymer based GML is if the GML contains a defect, e.g., a puncture hole or crack, or is otherwise defective due to poor manufacturing or poor placement and protection procedures. In such cases, the GML will offer essentially no resistance to contaminant migration through the defect, such that contaminant migration will readily pass through the GML, i.e., unless the GML is founded upon a hydraulic resistant layer, such as natural, lowpermeability clay, or the GML represents the upper component of a composite liner which includes an underlying low-permeability component, such as a CCL or geosynthetic clay liner (GCL). In the case where the GML is entirely intact, the only way aqueous-phase contaminants can pass through the

Figure 7. Schematic of concentration profile for organic chemical diffusion through an intact geomembrane liner (GML) (modified after Rowe 1998, Katsumi et al. 2001).

Since GMLs are relatively thin, steady-state diffusion through the GML can be established relatively quickly, such that the mass flux of the organic chemical can be expressed in accordance with Fick's first law as follows (Park et al. 1996a,b, Rowe 1998, Katsumi et al. 2001, Rowe 2005):

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C  Ce J d  Dg K g o Lg

Geosynthetic clay liners (GCLs) are relatively new barrier materials, having first been used in a landfill in 1986 (Bonaparte et al. 2002). Traditional or conventional GCLs are thin (~ 5 to 10 mm), prefabricated (factory manufactured) hydraulic barriers (liners) that consist primarily of a processed clay, typically sodium bentonite, or other low permeability material that is either encased or "sandwiched" between two geotextiles or attached to a single polymer membrane (i.e., geomembrane) and held together by needle-punching, stitching, and/or gluing with an adhesive. The hydraulic resistance of these conventional GCLs that do not include a geomembrane or polymer film is attributed to the bentonite component of the GCL, which swells in the presence of water to form a tight sealing layer. Although GCLs can be subjected to significant incompatibility upon permeation with chemical solutions or liquids, resulting in potentially significant increases in hydraulic conductivity, the values of kh for GCLs permeated with dilute chemical solutions or water tend to be less than about 1 x 10-10 m/s (e.g., Shackelford et al. 2000). Such low kh values and the relative thinness of GCLs imply that diffusion would be a significant, if not dominant, transport process through GCLs. Accordingly, several studies have evaluated the diffusion of chemicals through GCLs (Lake and Rowe 2000, 2005, Rowe et al. 2000, Malusis and Shackelford 2002a, Lange et al. 2009, Paumier et al. 2011, Malusis et al. 2013). For example, consider the results of the study shown in Fig. 9 for diffusion of KCl through a GCL. In this study, diffusion of KCl was hypothesized to be affected by the ability of the bentonite in the GCL to exhibit semipermeable membrane behavior, whereby solutes are excluded from the smaller pores in the clays, thereby restricting the diffusion of the KCl (Malusis and Shackelford 2002b). Such solute restriction also results in chemico-osmosis, or the movement of liquid from lower solute concentration to higher solute concentration, or opposite to the direction of diffusion. Accordingly, the GCL was tested in an apparatus that was able to measure simultaneously both the membrane efficiency of the GCL and the D* of the KCl. The membrane efficiency refers to the relative degree or extent of solute restriction (also referred to as "ion exclusion"), and is quantified in terms of a membrane efficiency coefficient,  (Shackelford et al. 2003). Although negative values of  have been reported in some cases due to atypical circumstances resulting from processes such as "diffusion-osmosis" (Olsen et al. 1990),  values typically range from zero for clays exhibiting no membrane behavior and, therefore, no solute restriction, to unity (100 %) for "perfect" or "ideal" membranes that restrict the passage of all solutes. Because soils generally exhibit a range of pore sizes, some of the pores in clays may be restrictive whereas others are not. As a result, most natural soils that exhibit membrane behavior do so as "imperfect" or "non-ideal" membranes, such that 0 <  < 1 (Shackelford et al. 2003). In particular, bentonite has been shown to possess the potential for significant membrane behavior, such that the possible effect of membrane behavior on solute transport through any bentonite-based barrier should be considered (Shackelford 2011, 2012, 2013). In terms of the results in Fig. 9, Fig. 9a shows the correlation between the measured value of  for the GCL and the source concentration of KCl, Co, used in the test. Due to physico-chemical interactions between the salts in the pore water of the bentonite and the bentonite particles,

(6)

where Dg is the diffusion coefficient for the chemical in the GML. Conservative (high) estimates of Jd will occur when Ce is assumed to be zero. Since geomembranes are not porous media, the nature of Dg is not the same as that of D*. For example, based on an extensive summary of both Kg and Dg values from the literature reported by Rowe (1998), the upper limit on the vast majority of the Dg values is on the order of 1 x 10-1l m2/s, with numerous values ranging from one to several orders of magnitude lower than this value. Thus, values of Dg generally are several orders of magnitude lower than values of D*. However, despite such low magnitude Dg values, Park et al. (1996b) illustrate that molecular diffusion of organic chemicals through intact GMLs can be substantially greater than leakage through geomembrane defects. A major reason for this difference is that that cross-sectional area for diffusive mass flux through a GML is the entire surface of the GML, whereas mass flux due to leakage through a GML is associated with only a small percentage of the surface area (see Fig. 8). Area, A

Diffusion

Leakage through Defect Area, Ad

Figure 8. Cross-sectional areas for diffusion versus leakage through a GML.

Because diffusion of VOCs through single polymer GMLs has been an issue, recent research has focused on evaluating alternative GMLs for the ability to minimize VOC diffusion. For example, Sangam and Rowe (2005) evaluated the effect of fluorinating the surface of an HDPE on the diffusion of VOCs through the GML. In essence, the surface fluorination reduces the affinity of the GML to VOCs. Sangam and Rowe (2005) reported that the diffusion coefficient for the surface fluorinated HDPE was on the order of 1.5 to 4.5 times lower than that for the untreated HDPE, depending on the specific hydrocarbon evaluated. Similarly, McWatters and Rowe (2010) evaluated the ability of two coextruded GMLs to reduce the diffusive flux of VOCs. Coextrusion involves extruding two or more layers of dissimilar polymers into a single film. McWatters and Rowe (2010) reported improved resistance to BTEX diffusion for the two coextruded GMLs, a polyamide (nylon) GML and an ethylene vinyl-alcohol (EVOH) GML, relative to that for either an LLDPE or a PVC GML. The results of these and other studies indicate that alternatives to the single polymer GMLs may offer improved performance in terms of VOC diffusion. 4.2.4

Diffusion through Geosynthetic Clay Liners

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Co for the granular bentonite used by Lake and Rowe (2000) may not be the same as that shown in Fig. 10 due, in part, to the different porosity of the specimens (n = 0.78 to 0.80 vs. n = 0.71), different salts used in the tests (KCl versus NaCl), and the potentially different properties of the granular bentonites in the two GCLs. Despite these differences, the results shown in Fig. 10 suggest that there is general agreement between the results reported in the two studies.

Membrane Efficiency Coefficient, 

1 0.8

Effective Diffusion Coefficient, D* (x 10 -10 m2/s)

higher salt concentrations result in compression of the adsorbed layers of cations associated with the bentonite particles and, therefore, larger pore openings between adjacent particles and lower . As shown in Fig. 9b, such larger pores due to higher salt concentrations also result in increasing values of D* for KCl with increasing Co. Note that the values of D* shown in Fig. 9b are steady-state values in that the values correspond to after steady-state diffusion had been established with respect to both Cl- and K+. The combined effect of Co on  and D* is shown in Fig. 9c, where D* is shown to decrease with increasing  such that, in the limit as  → 1, D* → 0 as required on the basis of the definition of a perfect or ideal membrane. As indicated in Fig. 9b, this decrease in D* with increasing  was attributed to a decrease in the apparent tortuosity factor, a (see Eq. 1).  = -0.457 - 0.455log(Co) (a) (r2=0.998)

0.6 0.4 0.2

Lake & Rowe (2000) Malusis & Shackelford (2002a)

5

Membrane Behavior (0 <  < 1)

4 3

KCl (n = 0.78 - 0.8)

2 1 0 0.001

NaCl (n = 0.71) No Membrane Behavior ( = 0)

0.01 0.1 1 Source Salt Concentration, Co (M)

10

Figure 10. Comparison of the results for the diffusion of salts through GCLs from two different studies (modified after Malusis and Shackelford 2002a).

3 D* (x 10-10 m2 /s)

(b)

4.2.5 Diffusion through Composite Liners Composite liners refer to engineered barriers that are comprised of more than one type of barrier in intimate contact with each other. Although there are a variety of possible composite liner systems, including those that contain more than two component types of barriers (e.g., Nguyen et al. 2011), the most common types of composite liners consist of a GML overlying and in intimate contact with either an underlying CCL or an underlying GCL, although other composite liner scenarios are possible. For these common composite liners, the effectiveness of the composite liner in restricting contaminant migration relies largely on the integrity of the overlying GML and on the intimacy of the contact between the overlying GML relative to the underlying CCL or GML (Rowe 1998, Foose et al. 2001, 2002). The fewer the number of defects in the GML and the more intimate (tighter) the contact between the two barriers, the more effective the barrier in restricting contaminant migration. However, failure to protect the GML could compromise the integrity of the composite liner. For example, Rowe et al. (2003) evaluated the performance of a composite liner comprised of a 1.5-mmthick HDPE GML overlying a 3-m-thick CCL after 14 years in operation as a leachate lagoon liner (also see Rowe 2005). The GML had been poorly protected, resulting in development of 528 defects (cracks, holes, patches) per hectare over the 14-yr operational life of the liner, which allowed leachate to seep between the GML and CCL. Data obtained upon decommissioning indicated that leachate leaking through the GML had spread quickly over the entire interface between the GML and CCL, essentially rendering the GML ineffective. However, there were questions as to when the GML became ineffective as a barrier component and to what extent contaminant had penetrated the underlying CCL. Based on these considerations, Rowe et al. (2003) evaluated the chloride concentration profile within the CCL based on samples recovered from five different locations. As illustrated in

2

1

0 0.001 0.01 0.1 Source KCl Concentration, Co (M)

3 D* (x 10 -10 m2/s)

D* = 2.4 x 10 -10 m 2/s

(c)

0.2

a,max = 0.12

2

0.1 1

*

D

a 0

0 0 0.2 0.4 0.6 0.8 1 Membrane Efficiency Coefficient, 

Apparent Tortuosity Factor, a

Effective Diffusion Coefficient,

Effective Diffusion Coefficient,

0 0.001 0.01 0.1 Source KCl Concentration, Co (M)

6

Figure 9. Results of a test to measure simultaneously the diffusion of KCl through a GCL and the membrane behavior of the GCL: (a) membrane efficiency of the GCL; (b) steady-state diffusion coefficient of KCl; (c) effect of membrane behavior on steadystate diffusion of KCl (modified after Malusis and Shackelford 2002a,b).

Malusis and Shackelford (2002a) compared their results with those reported by Lake and Rowe (2000) based on measurement of NaCl diffusion under constant volume conditions through granular sodium bentonite extracted from a GCL. The results of this comparison are shown in Fig.10 in the form of the D* values for KCl and NaCl versus the source salt concentration, Co. Overall, results in Fig. 10 indicate a similar trend of increasing D* with increasing Co. Although  values were not measured by Lake and Rowe (2000), chemico-osmotic flow was reported to be sufficiently negligible such that the authors concluded that membrane behavior probably wasn't significant for the range of NaCl concentrations used (i.e., Co ≥ 0.08 M). The superimposed demarcation between membrane behavior ( > 0) and no membrane behavior ( = 0) based on the results shown in Fig. 9 tends to support this conclusion, although the relationship between  and 10

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Fig. 11, the resulting concentration profile was consistent with diffusion dominated transport, taking into consideration that reverse or back diffusion had occurred at the top of the profile due to the placement of water within the lagoon following removal of the leachate prior to decommissioning, resulting in a localized reversal in the concentration gradient. Additional calculations were performed to evaluate the duration of the effectiveness (i.e., lifespan) of the GML on the resulting concentration profiles, with the results indicating that the GML likely was effective only for an initial period ranging from 0 to 4 yr. Nonetheless, the overall conclusion was that diffusion was the dominant transport process, and the underlying groundwater was not impacted due to the 3-m thickness of the CCL.

0

(ANOVA) of the measured concentrations, Klett (2006) concluded that the concentrations for 8 of the 11 VOCs were statistically no different between clay and composite lined landfills. Containment Liquid (Co) GML CCL or GCL

Diffusion (a) Upgradient Side

Chloride Concentration (mg/L) 200 300 400 500

Transient Profiles

Depth Below GML (m)

Lifespan = 8 yr Lifespan = 0.5 10 yr

(0-,t) = Co C(0+,t) = KgCo

GML

1

1.5

(Lg+,t) = Ci CCL or GCL

CCL Parameters: *

-10

m /s

-10

m/s

D = 7 x 10

2

(Lg-,t) = KgCi

Lifespan = 0 yr Lifespan = 4 yr Lifespan = 6 yr

Measured

kh = 2 x 10

2

(Lg+Lc,t) = Ce Downgradient Side

n = 0.45 @ 0.0-0.25 m e

(b)

= 0.42 @ 0.25-0.5 m = 0.38 > 0.5 m

Figure 12. Schematics of diffusion of VOCs through intact composite liners: (a) conceptual transport; (b) concentration profiles (modified after Foose et al. 2001, Foose 2002)

2.5

Figure 11. Measured and predicted chloride concentration profiles through the compacted clay portion of a composite liner system after 14 yr of operation (modified from Rowe et al. 2003, Rowe 2005).

Although there is substantial evidence indicating that composite liners are effective in terms of waste containment, i.e., when constructed properly (e.g., Sangam and Rowe 2001b, Bonaparte et al. 2002, Rowe 2005), there also is growing evidence the composite liners are not any more effective against minimizing VOC transport than are CCLs (e.g., Foose 2002, Foose et al. 2002, Shackelford 2005, Klett 2006). In this case, the VOC first must diffuse through the overlying GML similar to the situation for the single GMLs illustrated in Fig. 7. However, once the VOC has partitioned out from the downgradient side of the GML, the VOC then must diffuse through the underlying CCL or GCL, as illustrated in Fig. 12. For example, Klett (2006) evaluated the measured concentrations of 11 VOCs existing in 94 lysimeters (e.g., Fig. 13) at 34 landfills in Wisconsin lined with either CCLs or composite liners (some landfills had multiple cells, each with a lysimeter). The lysimeter data set consisted of 2738 samples analyzed for VOCs. At least one VOC with a concentration above the limit of detection was detected in 1356 of these samples, and at least one VOC was detected during one sampling event in each of the 94 lysimeters evaluated. Toluene was detected most frequently (60% of the lysimeters) and ten VOCs (toluene, tetrahydrofuran, dichloromethane, benzene, acetone, chloromethane, xylene (total), ethylbenzene, trichloroethylene, and 1,1dichloroethane) were detected in more than 25 % of the lysimeters. The most prevalent compounds were aromatic hydrocarbons (toluene and benzene), furans (tetrahydrofuran), and the alkanes (dichloromethane and 1,1- dichloroethane). Based on analysis of variance

Figure 13. Schematic of typical collection lysimeter (underdrain) beneath a composite liner for a solid waste disposal facility (modified from Shackelford 2005).

An example of this comparison for dichloromethane (DCM) is presented in the form of box plots shown in Fig. 14. The center line in each box plot represents the median of the data, the outer edges of each box represent the interquartile range (i.e., 25th to 75th percentiles), and the outermost lines or "whiskers" represent the 5th and 95th percentiles. As shown in Fig. 14, the concentrations of DCM in collection lysimeters beneath composite lined cells were not any lower than those collected beneath cells lined only with compacted clay. This similarity in DCM concentrations is not necessarily surprising, given that aforementioned lack of resistance to VOC diffusion offered by most geomembranes. Thus, diffusion of VOCs through GML-based composite liners remains an issue that must be addressed when such contaminants are present.

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4

Direction of Advection

3

Direction of Diffusion

+x

2 1 ES

0

>0

-1

=0

PAL

-2 Clay

Composite

Figure 14. Box plot comparisons of dichloromethane (DCM) concentrations in collection lysimeters beneath composite lined and clay lined cells in landfills in Wisconsin; ES = enforcement standard; PAL = protective action limit (data from Klett 2006).

4.2.6 Vertical Barriers A wide variety of vertical barriers have been used for in situ hydraulic and contaminant containment applications, including sheet-pile walls, grout curtains, concrete barriers, geomembrane barriers, gravel-filled trenches, and slurry based cutoff walls, such as soil-bentonite (SB), cementbentonite (CB) and soil-cement-bentonite (SCB) walls (Mitchell et al. 2007). However, the slurry based vertical cutoff walls probably are the most commonly used vertical barriers for in situ containment of contaminants. Similar to the case of horizontal barriers (Fig. 3), contaminant transport through such vertical barriers can be categorized into three possible scenarios as illustrated in Fig. 15, viz., pure diffusion (Fig. 15a), diffusion with positive (outward) advection (Fig. 15b), and diffusion with negative (inward) advection (Fig. 15c). The pure diffusion scenario (Fig. 15a) exists when there is no applied hydraulic gradient across the barrier. This scenario would exist only in practice when there was little or no local groundwater flow in the vicinity of the barrier location prior to installation of the barrier, and no net accumulation or depletion of water on either side of the barrier during the operational life of the barrier. As a result, the only possible transport process is diffusion from the containment (inward) side of the barrier (C > 0) towards the outside of the barrier (C = 0). As the conditions for this scenario are not typically encountered in practice, this scenario may be considered as a limiting case. The scenario for diffusion with positive (outward) advection (Fig. 15b) exists when the local groundwater level on the containment side of the barrier is allowed to rise, e.g., via infiltration of precipitation, such that a hydraulic gradient is established across the barrier in the same direction as the prevailing concentration gradient, i.e., from the containment (inward) side of the barrier (C > 0) towards the outside of the barrier (C = 0). Thus, both advection and diffusion occur in the same direction, i.e., outward. The scenario for diffusion with negative (inward) advection (Fig. 15c) is analogous to the hydraulic trap scenario represented in Fig. 3c, and occurs when the groundwater level within the containment side is drawn down, e.g., by pumping or passive drainage (e.g., French drains), so as to generate an inwardly directed hydraulic gradient to drive advective transport that counteracts the outwardly directed diffusive transport, thereby minimizing the net outward contaminant flux. Transport analyses for this scenario have been reported by Shackelford (1989), Manassero and Shackelford (1994), Devlin and Parker (1996), and Neville and Andrews (2006).

(a) Diffusion without Advection (Pure Diffusion)

>0

=0

(b) Diffusion with Positive Advection

>0

=0

(c) Diffusion with Negative Advection

Figure 15. Contaminant transport scenarios across vertical barriers for in situ containment: (a) pure diffusion; (b) diffusion with positive (outward) advection; (c) diffusion with negative (inward) advection (modified after Gray and Weber 1984, Shackelford 1989, 1993, Manassero and Shackelford 1994, Devlin and Parker 1996, Neville and Andrews 2006, Sleep et al. 2006, Mitchell et al. 2007).

Although several studies have focused on evaluating contaminant transport through slurry based vertical cutoff walls (Gray and Weber 1984, Mott and Weber 1991a,b, Manassero et al. 1995, Devlin and Parker 1996, Khandelwal et al. 1998, Rabideau and Khandelwahl 1998, Krol and Rowe 2004, Britton et al. 2005, Neville and Andrews 2006, Malusis et al. 2010), only a few of these studies (e.g., Mott and Weber 1991a,b, Khandelwal et al. 1998, Krol and Rowe 2004) were extensively experimental studies focusing specifically on evaluating the diffusive properties of contaminants in traditional (unamended) SB backfills. In all of these studies, which were focused on diffusion and sorption of organic chemicals (e.g., 1,4dichlorobenzne, 4-chlorophenol lindane, trichloroethylene, and aniline), the results indicated that the values of D* typically were reduced by a factor of only about two to four relative to the corresponding values of Do, and at most were no more than an order of magnitude lower than Do, due, in part, to the relative high porosity values associated with most SB backfills. Also, sorption of the organic chemicals to the traditional (unamended) soil-bentonite backfills typically was negligible (i.e., Kd  0) due to the typically low organic carbon contents of the unamended backfill materials (e.g., Malusis et al. 2010). These two factors (i.e., relatively high D* and negligible Kd) combined with the typical inability to achieve backfill hydraulic conductivity values lower than about 10-10 m/s (e.g., D'Appolonia 1980, Evans 1991, 1993, 1994, Filz and Mitchell 1996, Shackelford and Jefferis 2000, Filz et al. 2003), suggest that the significance of diffusive transport across vertical cutoff walls is likely governed largely by the magnitude of the applied hydraulic gradient, ih, across the barrier, with diffusive transport becoming more significant with decreasing magnitude in ih (i.e., Fig. 15a). 12 138

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Nonetheless, prudence dictates consideration of diffusive transport in terms of long-term performance assessments, as the results of several simplified transport analyses suggest that diffusion may be significant under some scenarios (e.g., Gray and Weber 1984, Shackelford 1989, Manassero and Shackelford 1994).

associated text). As a result of these advancements, and the continuing need to assess the performance of the containment structures used to isolate HLRW from the environment for extensive time frames, diffusion of radionuclides through bentonite buffer barriers is likely to remain an important research area for the foreseeable future.

4.2.7

Diffusion through Bentonite Buffers for HighLevel Radioactive Waste (HLRW) Disposal Diffusion of radionuclides through highly compacted bentonites being considered as buffer barriers in HLRW disposal scenarios has been an area of substantial research over the past several decades, and in particular the past approximate decade. In fact, the number of referenced publications focused on evaluating diffusion of radionuclides through bentonite buffer barriers for HLRW disposal is too voluminous to cite here, but a representative listing can be found in Shackelford and Moore (2013). The high number of publications in this area results from the need for safe and secure, long-term disposal of HLRW (e.g., ≥ 10,000 yr) resulting from the significant past and present roles of nuclear energy in several countries (e.g., Belgium, Canada, France, Japan, Spain, Switzerland, United Kingdom, and the USA). In particular, two issues related to radionuclide diffusion through highly compacted bentonite buffers have been identified, viz., the influence of surface and/or interlayer diffusion, and the existence of semipermeable membrane behavior as a result of ion exclusion (Shackelford and Moore 2013). Surface or interlayer diffusion refers to the diffusion of cations, typically metals, sorbed to clay particles in addition to diffusion of cations within the mobile pore water between particles, i.e., outside the extent of influence of the negative electrical potentials associated with the individual clay particle surfaces. This phenomenon is attributed to the excess of sorbed cations in the diffuse double layers surrounding negatively charged clay surfaces relative to the concentration of cations that exists in the mobile pore water, and is known as interlayer diffusion when referring to the excess of sorbed cations within the interlayer regions of smectitic based clays, such as bentonites (Glaus et al. 2007, Appelo et al. 2010). When prevalent, surface and/or interlayer diffusion can result in enhanced diffusion of cations, and diminished diffusion of anions, relative to the diffusion of neutral tracers such as tritium and deuterium (Appelo et al. 2010). However, Shackelford and Moore (2013) noted that conflicting results have been reported as to the significance of surface and/or interlayer diffusion, and that the phenomenon is likely to be significant only in high activity clays, such as bentonites, compacted at relatively high dry densities. Also, the significance of surface and/or interlayer diffusion will be a function of the chemical speciation of the diffusing radionuclide. In terms of semipermeable membrane behavior, numerous studies have reported significant ion exclusionary properties of bentonite buffer barriers, but these properties historically have been taken into account qualitatively or indirectly by incorporating a correction (anion exclusion) factor within the form of Fick's first law (Shackelford and Moore 2013). However, recent advances in simultaneously testing for both solute diffusion and semipermeable membrane behavior as previously documented for GCLs have largely eliminated this restriction, such that quantification of the effect of semipermeable membrane behavior of radionuclide diffusion can now be assessed (e.g., see Fig. 9 and

4.3

Diffusion as an Attenuation Mechanism (Matrix Diffusion)

The process of matrix diffusion, whereby contaminants diffuse from interconnected pores or fractures into the surrounding intact clay or rock matrix, may be an important attenuation mechanism when the contaminant transport occurs through structured clay and/or rock formations (e.g., Foster 1975, Grisak and Pickens 1980, Neretnieks 1980, Feenstra et al. 1984, Lever et al. 1985, Rowe and Booker 1990, 1991, Boving and Grathwohl 2001, Polak et al. 2002, Lipson et al. 2005). In this regard, matrix diffusion has been considered in terms of the migration of radionuclides resulting from high-level radioactive waste disposal through fractured crystalline rocks (Neretnieks 1980, Sato 1999), the migration of pesticides resulting from agricultural practice through fractured clayey till (Jorgensen and Fredericia 1992, Jorgensen and Foged 1994), the migration of leachate resulting from solid waste landfills through underlying fractured clayey till (Rowe and Booker 1990, 1991), and the migration of dense-chlorinated solvents resulting from industrial spills and disposal practice through fractured geologic media (Parker and McWhorter 1994, Parker et al. 1994, 1996). For example, consider the scenario depicted in Fig. 16 after Rowe and Booker (1990, 1991), whereby a clay-lined (CCL) waste containment facility is underlain by fractured till that serves as an "attenuation layer" (AL) that could attenuate the migration of any contaminants emanating from the containment facility to the underlying confined aquifer. In this scenario, the greater the ability of the fractured till to attenuate the migration contaminants, the more effective the overall or global containment system (i.e., CCL + AL). In this regard, the fractures may serve as conduits that facilitate the rate of downward migration of contaminants, but matrix diffusion of contaminants from the fractures into the surrounding intact clay matrix and any subsequent sorption of the contaminants to the individual clay particles within the matrix pores can provide for an effective retardation of advancing, downward contaminant migration. Matrix diffusion also may be important in attenuating the migration of contaminants at the local or barrier scale. For example, Jo et al. (2006) proposed a threecompartment model that included rate-limited cation exchange controlled by matrix diffusion to explain the extensive tailing of eluted cations that often is observed during column tests conducted on aggregated soils with inorganic chemical solutions. As illustrated schematically in Fig. 17, the pore space in the saturated granular bentonite was assumed to consist of intergranular, interparticle, and interlayer (interlaminar) spaces. The pores between the granules constituted the intergranular pore space, whereas the interparticle pore spaces existed between the particles comprising the granules, but outside the interlayer space between the montmorillonite lamella. Water in the intergranular pore space was assumed to be hydraulically mobile. Water in the interparticle and 13

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

interlayer pores was assumed to be strongly bound by electrostatic forces and immobile. Ion exchange was assumed to occur as cations in the permeant solution passed through the intergranular pores (Fig. 17a) and gradually diffuse first into the interparticle pores (Fig. 17b) and subsequently into interlayer spaces (Fig. 17c). Cation exchange progressed until equilibrium was established between cations in the permeant solution and the montmorillonite surface.

Figure 17. Effect of diffusion on solute migration through a GCL containing granular bentonite (modified after Jo et al. 2006 and Shackelford and Moore 2013).

Ca Effluent Concentration (mmol/L)

Figure 16. Idealized schematic cross section of lined waste containment system underline by fractured till and the concept of attenuation via lateral diffusion from fracture into the intact surrounding till matrix (modified after Rowe and Booker 1991).

Jo et al. (2006) compared measured breakthrough curves (BTCs) for calcium (Ca) transport through specimens of a GCL based on the results of six column tests versus predicted BTCs based on their theoretical model. The results of this comparison are shown in Fig. 18, wheregp (s–1) is the mass transfer coefficient for diffusion between the mobile intergranular and the immobile interparticle liquids, and pl (s–1) is the mass transfer coefficient for diffusion between the immobile interparticle and interlayer liquids. The predictions obtained with the model for the base case generally were comparable to the data, even though the model input parameters were estimated independently (i.e., the parameters were not determined from calibration). The model also predicted reasonably well the changes in the exchange complex, but the comparison between the predicted and measured eluted sodium (Na) concentrations was not quite as favorable (see Jo et al. 2006). Nonetheless, the results of the study by Jo et al. (2006) serve as an example of the role that diffusion can play as an attenuation mechanism during solute transport through barriers comprised of structured soils.

25 

gp

=  = 6.0x10 s -8

-1

pl

20 15 

gp

10

-5

-1

-7

-1

= 4.0x10 s

 = 4.0x10 s pl

5 

=  = 4.0x10 s -5

gp

-1

pl

Test 1 Test 2 Test 3 Test 4 Test 5 Test 6 Predicted High  Low 

0 0

20

40 60 80 Pore Volumes of Flow

100

120

Figure 18. Measured and predicted breakthrough curves for calcium (Ca) transport through a GCL containing granular bentonite where matrix diffusion plays a significant role as an attenuation mechanism (modified after Jo et al. 2006).

4.4 Liquid-Phase Diffusion in Unsaturated Media Although the vast majority of studies have focused on liquid-phase diffusion of chemicals through saturated porous media, there are a wide variety of applications in environmental geotechnics where liquid-phase diffusion through unsaturated porous media can be an important consideration. Some of the possible applications include diffusion of salts through unsaturated layers within an engineered cover system and the potential impact of such salts on the integrity of GCLs used as a component of the cover system (e.g., Benson and Meer 2009, Scalia and Benson 2011, Bradshaw et al. 2013), diffusion of radionuclides through unsaturated coarse-grained layers surrounding subsurface radioactive and hazardous waste 14

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Cover

repositories (Conca and Wright 1990), and unsaturated diffusion of chemicals in the vadose zone (Charbeneau and Daniel 1993). As an example of this last scenario, Rowe and Badv (1996b) evaluated the diffusion of chloride and sodium through a layered soil system consisting of an essentially saturated clayey silt overlying an unsaturated layer of either coarse sand or fine gravel. This two-layered soil system mimics the practical situation where a landfill may be sited in a hydrogeological setting where the predominant soil type below the proposed landfill base is granular (e.g., silt, sand, or gravel) and the water table is also at some depth. In this case, Sw of the soil below a liner may be expected to increase from about residual w below the liner to almost full saturation (Sw = 1) at or near the water table. Based on the results of their study, Rowe and Badv (1996b) found that the value of D* for chloride and sodium in the unsaturated soil, or D*unsat, relative to that in the saturated soil, D*sat, could be approximated reasonably well by a simple linear function of w, or D*unsat/D*sat = w/n. Also, Rowe and Badv (1996b) concluded that, provided that the Darcy velocity can be kept low (e.g., by the construction of a good compacted clay or composite liner), the unsaturated fine gravel evaluated in their study may act as a diffusion barrier to the migration of the dissolved sodium and chloride ions.

H2O

Tailings Dam

O2

Sulphidic Tailings

Tailings Dam

Acid Drainage (pH < 2)

(a) Cover

Tailings Dam

Radon

H 2O

Uranium Tailings

Tailings Dam

Radionulcides (b)

Figure 19. Tailings disposal scenarios where gas-phase diffusion plays an important role: (a) oxidation of sulphidic tailings and generation of acid drainage; (b) radon gas emission (modified after Shackelford and Nelson 1996, Shackelford 1997).

For example, Stormont et al. (1996) evaluated the effect of unsaturated flow through the three cover sections shown in Fig. 20a in terms of the effective air-phase diffusion coefficient for oxygen gas (O2(g)), De, at a depth of 0.6 m (i.e., the interface between the cover and the underlying material). Their results are shown in Fig. 20b in the form of a normalized oxygen diffusion coefficient, DN, defined as follows (e.g., see Charbeneau and Daniel 1993, Stormont et al. 1996):

4.5 Gas-Phase Diffusion Gas-phase diffusion can be an important consideration in environmental geotechnics, including both waste containment applications (e.g., Yanful 1993, Aubertin et al. 2000, Mbonimpa et al. 2003, Aachib et al. 2004, Bouzza and Rahman 2004, 2007, Alonso et al. 2006, Demers et al. 2009) and remediation applications, such as in the use of the soil vapor extraction technology for removal of VOCs from the subsurface vadose zone (e.g., Johnson et al. 1990). The importance of gas-phase diffusion is accentuated because diffusion coefficients for chemicals in the gas-phase typically are four-to-five orders of magnitude greater than those for the same chemicals in the liquid phase (Cussler 1997). For this reason, the gasphase diffusive mass flux of a chemical through soil can be reduced significantly by minimizing the continuity in the gas (air) phase of the medium, for example, by filling the voids with a sufficient amount of water such that the gas phase becomes discontinuous (e.g., Nicholson et al. 1989, Yanful 1993, Bouzza and Rahman 2004, 2007). Two waste containment problems of interest involving gas-phase diffusion and the environmentally safe disposal of mine tailings are illustrated schematically in Fig. 19. The problem of acid drainage (Fig. 19a) occurs when sulphidic tailings (e.g., pyrite or FeS2) are oxidized resulting in the production of a low pH solution (e.g., pH ≈ 2) that leaches potentially toxic heavy metals associated with the tailings during percolation through the tailings, resulting in the emanation of acid drainage form the tailings (e.g., Nicholson et al. 1989, Evangelou and Zhang 1995, Ribet et al. 1995). In the case of the disposal of uranium tailings (Fig. 19b), the tailings can serve as a localized source of radon gas that can be environmentally harmful if not controlled properly. In both of these cases, the objective in the cover design must include steps taken to minimize diffusive influx (O2) or diffusive efflux (radon) of gas through the cover.

 DN

10/3

De  a   De,max  n 

(7)

where De is the effective air-phase diffusion coefficient (= aaDa), a is the volumetric air content, n is total soil porosity, a is the apparent tortuosity factor for the air phase (= a7/3/n2), Da is the pure air-phase diffusion coefficient ( 2.26 x 10-5 m2/s), and De,max = De at a = n. Thus, 0 ≤ DN ≤ 1, such that diffusion of O2(g) via the air phase will be minimized as a approaches zero (a → 0). However, as shown by Aachib et al. (2004), minimizing the diffusion of O2(g) via the air-phase does not necessarily mean that the liquid-phase diffusion of O2(g) also will be unimportant. As shown in Fig. 20b, DN for the monolithic and resistive covers remained relatively high because the water content at the 0.6-m depth tended to remain relatively constant at the field capacity of the soils. However, in the case of the capillary barrier, DN was significantly lower and more variable than the other cover sections, because the water content immediately above the interface between the finer and coarser layers remained high due to the capillary barrier effect. Stormont et al. (1996) attributed the variability in DN to the variability in water contents associated with wet and dry seasons.

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Capillary Barrier 60 mm

relatively quickly, resulting in a reversal of the concentration gradient and an outward diffusive flux of the contaminant (Fig. 21b). This outward or reverse matrix (back) diffusion process results in a slow release of residual contamination back into the aquifer that can lead to failure of the pump-and-treat remediation technology to achieve regulatory levels within a short time frame, leading to extensive pumping and excessive costs (e.g., Feenstra et al. 1996).

Resistive Cover

Monolithic Cover

15 mm

60 mm

45 mm

30 mm Gravel (k = 0.1 m/s) Uncompacted Soil (k = 1.4 x 10-6 m/s) Compacted Soil (k = 6.9 x 10-8 m/s) (a) -1 -2

10

-3

10

N

Normalized Oxygen Diffusion Coefficient, D

10

-4

10

-5

10

-6

10

Capillary Barrier Monolithic Cover Resistive Cover

-7

10

(b)

-8

10

0

100

200 300 Time (d)

400

Figure 20. Gas-phase oxygen diffusion through three types of soil covers: (a) cross sections of cover types; (b) normalized oxygen diffusion coefficients at 0.6-m depths within the soil covers (data from Stormont et al. 1996; modified after Shackelford 1997).

Figure 21. Matrix diffusion and reverse matrix diffusion: (a) diffusion into clay lens before pump-and-treat remediation; (b) reverse matrix or back diffusion out of contaminated clay lens during pump-and-treat. (modified after Shackelford and Lee 2005).

5 DIFFUSION IN REMEDIATION APPLICATIONS In terms of remediation, failure of the pump-and-treat technology to achieve clean-up goals has been attributed, in part, to the process of "reverse matrix" or "back" diffusion resulting in the slow and continuous release of contaminants from the intact clay and rock matrix into the surrounding, more permeable media, such as fractures or aquifer materials (e.g., Mackay and Cherry 1989, Mott 1992, Feenstra et al. 1996, Shackelford and Jefferis 2000, Chapman and Parker 2005, Seyedabbasi et al. 2012). Diffusion also has long been recognized as the transport process that controls the potential leaching of contaminants from stabilized or solidified hazardous waste, typically by the addition of pozzolanic materials such as cement, lime, and fly ash (e.g., Nathwani and Phillips 1980). Finally, diffusion may be a significant transport process with respect to controlling the rate of delivery of chemical oxidants (e.g., potassium permanganate, KMnO4) injected into contaminated low-permeability media through hydraulic fractures for in situ treatment of chlorinated solvents (Siegrist et al. 1999, Struse et al. 2002).

The effect of matrix diffusion on pump-and-treat remediation can be analyzed via superposition of an analytical solution based on the analogy between consolidation and diffusion and the principle of superposition (Shackelford and Lee 2005). For example, consider the case where the aquifer is initially contaminated with trichloroethylene (TCE) at a concentration, Co, of 1000 ppm, such that TCE diffuses into a 1-m-thick (= H) clay lens for a period of time. However, before the clay lens becomes completely contaminated, pump-and-treat remediation is undertaken to clean up the aquifer. As a result, the initial TCE concentration profile within the 1-m-thick (= H) clay lens is sinusoidal as a result of incomplete matrix diffusion of TCE into the clay lens prior to pumping, with a maximum TCE concentration of 1000 ppm at the aquifer-clay interface and a minimum contaminant concentration of 300 ppm at the center of the clay lens. This initial distribution of contaminant within the clay lens is represented in Fig. 22a in terms of the relative concentration, C(Z,T*)/Co, of TCE as a function of the dimensionless depth, Z, corresponding to a value of the dimensionless diffusive time factor, T*, of zero (T* = 0), where (Shackelford and Lee 2005):

5.1 Reverse Matrix or Back Diffusion As an example of reverse matrix or back diffusion, consider the scenario illustrated conceptually in Fig. 21a, where initial contamination of the aquifer results in a difference in concentration between the contaminated aquifer and the clay lens resulting in diffusion of contaminants into the porous matrix of the clay lens. After pumping commences, the higher permeability portion of the heterogeneous aquifer is flushed of contamination

Da t z D*t (8)  ; T*  Hd Rd H d 2 H d 2 and Hd is the maximum diffusive distance (= H/2 or 0.5 m in this example). The definition for the dimensionless Z 

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(i.e., U* = 0.90) corresponds to T* of approximately 0.76, which is equivalent to 95 yr. Thus, this analysis indicates that approximately a century of pumping would be required to remove 90 % of the initial contaminant mass from a partially contaminated, 1-m-thick clay lens, which is consistent with the aforementioned observations attributing failure of some pump-and-treat systems to reverse matrix diffusion. A similar analysis was presented by Feenstra et al (1996), but they assumed that the clay lens was initially completely (i.e., uniformly) contaminated. Such complete contamination of non-fissured clay lenses via matrix diffusion would be likely only in the case of relatively thin clay lenses and/or relatively long durations of aquifer contamination. Otherwise, the clay lenses likely would only be partially contaminated resulting in an initial concentration distribution within the clay lenses that is sinusoidal, thereby requiring the need for superposition in the resulting analysis (Shackelford and Lee 2005). Regardless of the level of contamination or the type of analysis required, this example supports the numerous observations that reverse matrix or back diffusion can play a significant role in affecting the remediation of contaminated aquifers.

depth is identical to that for the case of consolidation, where Hd is the maximum drainage distance, whereas the definition for the diffusive time factor, T*, is identical to that for the dimensional consolidation time factor, T, where Da is replaced by the coefficient of consolidation, cv (Shackelford and Lee 2005). On the basis that pumping results in "instantaneous" removal of contaminant from the surrounding aquifer at time t = T*= 0, the resulting contaminant concentration profiles for T* > 0 can be determined by means of superposition as shown in Fig. 22a. At times, T*, less than about 0.1, both outward diffusion at the boundaries and inward diffusion near the center of the clay lens are occurring simultaneously, whereas after T*  0.1, the concentration profiles have dissipated to the extent that only outward diffusion of TCE occurs. The dissipation of residual contamination will proceed over time until all of the contaminant initially within the clay lens has diffused into the surrounding aquifer and been removed. However, this mass removal can take considerable time. 0

0.001 (a)

Dimensionless Depth, Z = z/H

0.01 0.03

0.1

0.5

T*=

0

0.3 0.2 1.0 1

5.2

0.4

0.7  0.5

1.5

Contaminated, subaqueous sediments represent a major environmental issue worldwide. One approach for dealing with this issue is to cap the sediments in situ. The caps should perform one or more of the following functions (Alshawabkeh et al. 2005): (a) physical isolation of the sediment; (b) sediment stabilization, in terms of preventing erosion and resuspension; and (c) reduction of dissolved contaminant flux. A conceptual schematic of the role of capping in situ sediments is illustrated in Fig. 23. Placement of the capping layer will reduce contaminant flux by (1) eliminating the bioturbation zone (i.e., mixing or dispersion caused by benthic organisms at the top several centimeters of the contaminated sediments), (2) increasing the length through which contaminants must migrate via advection and diffusion, (3) retarding contaminant migration via sorption to the capping materials, and (4) eliminating resuspension and direct desorption of contaminants to the overlying water column (Wang et al. 1991, Thoma et al. 1993). Capping materials do not necessarily have to be low permeability soils, as typically is the case with covers for above ground waste disposal, but the materials should possess some sorption capacity to minimize the rate of contaminant migration through the cap. Initially, contaminant migration through the cap will occur both via advection and diffusion. The advective component of contaminant transport results from generation of excess pore-water pressures within the contaminated sediments due to placement of the capping material and the associated sediment consolidation. Some studies have indicated that consolidation induced contaminant mass flux can be several times greater than that due to diffusion during the initial, transient period when consolidation of the sediments is pronounced (e.g., Alshawabkeh et al. 2005). Nonetheless, diffusion still may play a significant role in terms of the contaminant mass flux through the cap during the initial transient transport stage of the process, and likely will be the dominant transport process under long-term, steady-state conditions

0.05 0.02 0.005

2 0

0.2

0.4

0.6

0.8

1

Relative Solute Concentration,C(Z,T*)/Co

Average Degree of Diffusion, U* (%)

0 (b) 20

40

60

80

100 0

0.5

1 T* ~ 0.76

1.5

Diffusion through Subaqueous Caps for Contaminated Sediments

2

Dimensionless Diffusive Time Factor,T*

Figure 22. Results of example analysis for the reverse matrix diffusion from 1-m-thick clay lens contaminated with TCE resulting in an initial sinusoidal contaminant distribution: (a) timedependency of relative TCE concentration versus dimensionless depth; (b) time dependency of the average degree of diffusion for contaminant removal (modified after Shackelford and Lee 2005).

For example, if we assume an Rd of 5.2 and D* of 3.33 x 10-10 m2/s for TCE and the clay based on Parker et al. (1996), then the degree of diffusion, U*, which represents the relative degree of mass removal (Shackelford and Lee 2005), 10 yr after the beginning of pumping (i.e., T* ~ 0.081) is approximately 0.43 or 43 %. Based on a porosity, n, of 0.60 for the clay lens and assuming complete reversibility of the sorbed TCE, the cumulative contaminant mass removed per unit area of the clay lens after 10 yr of pumping is approximately 749 g/m2 (Shackelford and Lee 2005). More importantly, as indicated in Fig. 22b, 90 % contaminant mass removal 17

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(Thoma et al. 1993). Although several studies have been undertaken towards the development of models that can handle the combined advective and diffusive mass transport of consolidating contaminated media (e.g., Smith 2000, Peters and Smith 2002, Alshawabkeh et al. 2005, Alshawabkeh and Rahbar 2006, Fox 2007a,b, Fox and Lee 2008, Lee and Fox 2009), comparatively fewer experimental studies for this scenario have been undertaken (e.g., Wang et al. 1991, Tang et al. 2005, Lee et al. 2009, Meric et al. 2010). Nonetheless, the issue of contaminant migration including diffusion from consolidating contaminated porous media remains an important area of research (e.g., Fox and Shackelford 2010).

underlying CCL or GCL. In this regard, relatively recent data from clay-lined and composite-lined landfills in Wisconsin, USA, indicate that the GML component of composite liners offers virtually no added resistance to VOC diffusion relative to the CCLs. Diffusion also may be an important consideration for contaminant transport through slurry based vertical cutoff walls, but the significance of diffusion in this case likely is governed by the magnitude of the hydraulic gradient, ih, across the wall, with the significance of diffusion increasing with decreasing ih. Finally, diffusion through subaqueous caps used for in situ containment of contaminated dredged sediments has been an area of significant study, particularly in terms of long-term environmental impacts. Matrix diffusion, whereby contaminants diffuse from interconnected pores or fractures into the surrounding intact clay or rock matrix, can be an important attenuation mechanism in assessing the potential environmental impact of migrating contaminants, both on a global scale such as beneath a landfill located over fissured or fractured clay or rock, and on a local scale such as through a GCL comprised of granular bentonite. However, the resulting contamination of the clay or rock matrix may result in ineffective and/or prolonged remediation of the sites due to the process of reverse matrix or back diffusion. Finally, gas-phase diffusion also can play a significant role in environmental geotechnics, particularly since diffusion via the gas phase can be significantly faster than that via the liquid phase. Two examples where gas-phase diffusion is important include the diffusion of oxygen through covers resulting in oxidation of sulphidic bearing mine tailings and the subsequent generation of acid drainage, and the release of radon from uranium bearing tailings to the surrounding atmosphere.

Air Water

Advection + Diffusion

Capping Layer

Contaminated Sediments

Figure 23. Schematic scenario of subaqueous cap for isolating contaminated sediments in situ.

6 CONCLUSIONS The role of diffusion in environmental geotechnics was reviewed. Diffusion has been shown to be a significant contaminant transport process through low-permeability barrier materials, including natural and engineered clay barriers such as compacted clay liners (CCLs) and geosynthetic clay liners (GCLs), with values of hydraulic conductivity, kh, lower than 10-9 m/s, and a dominant transport process for kh values lower than about 2-5 x 10-10 m/s. The increasing significance of diffusion with decreasing kh results in a situation whereby design of engineered clay barriers solely based on achieving low kh is not only incorrect but also unconservative with respect to the duration of contaminant containment in such situations. As a result, achieving low kh is a necessary, but not sufficient condition for assuring effective containment of contaminants with low- kh barriers. The existence of semipermeable membrane behavior is shown to affect the diffusion of simple salt solutions through bentonite based GCLs via ion exclusion. The greater the magnitude of the membrane behavior, the lower the effective diffusion coefficient. However, membrane behavior also is shown to diminish with increasing salt concentration, such that membrane behavior likely will play a minor, if any, role in affecting solute diffusion through traditional sodium bentonite based GCLs in many practical applications, such as landfills. Nonetheless, membrane behavior is likely to play a more significant role in terms of the diffusion of contaminants through other types of bentonite based barriers. For example, semipermeable membrane behavior is likely to be important in assessing diffusion of radionuclides through the highly compacted bentonite buffers being considered for containment of high-level radioactive waste, especially given the extremely long containment durations (e.g., 10,000 yr) associated with this application. Diffusion is known to be the dominant liquid-phase transport process of VOCs through intact geomembrane liners (GMLs), either alone or as a component of a composite liner overlying and in intimate contact with an

7 ACKNOWLEDGMENTS The author expresses his sincere gratitude to David Daniel, R. Kerry Rowe, Robert Quigley (deceased), John Cherry, Robert Gillham, and Donald Gray for their guidance and assistance during his PhD graduate studies on the topic of diffusion through clay barriers. The author also thanks Mario Manassero, Chair of ISSMGE Technical Committee TC215 on Environmental Geotechnics, for his support in receipt of the first R. Kerry Rowe Honorary Lecture which served as the basis for this paper. Finally, the author appreciates the assistance of his Ph.D. graduate student, Kristin Sample-Lord, in the preparation of this paper. 8 REFERENCES Aachib, M., Mbonimpa, M., and Aubertin, M. 2004. Measurement and prediction of the oxygen coefficient in unsaturated media, with applications to soil covers. Water, Air, and Soil Pollution, 156 (1-4), 163-193. Alonso, E.E., Olivella, S., and Arnedo, D. 2006. Mechanisms of gas transport through clay barriers. Journal of Iberian Geology, 32 (2), 175-196. Alshawabkeh, A.N., and Rahbar, N. 2006. Parametric study of one-dimensional solute transport in deformable porous media. Journal of Geotechnical and Geoenvironmental Engineering, 132 (8), 1001-1010. Alshawabkeh, A.N., Rahbar, N., and Sheahan, T. 2005. A model for contaminant mass flux in capped sediment under consolidation. Journal of Contaminant Hydrology, 78 (3), 147-165.

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Appelo, C.A.J., Vinsot, A., Mettler, S., and Wechner, S. 2008. Obtaining the porewater composition of a clay rock by modeling the in- and out-diffusion of anions and cations from an in situ experiment. Journal of Contaminant Hydrology, 101 (1-4), 67-76. Appelo, C.A.J., Van Loon, L.R., Wersin, P. 2010. Multicomponent diffusion of a suite of tracers (HTO, Cl, Br, I, Na, Sr, Cs) in a single sample of Opalinus Clay. Geochimica et Cosmochimica Acta, 74 (4), 1201-1219. Aubertin, M., Aachib, M., and Authier, K. 2000. Evaluation of diffusive gas flux through covers with a GCL. Geotextiles & Geomembranes, 18 (2-4), 215233. Badv, K. and Abdolalizadeh, R. 2004. A laboratory investigation on the hydraulic trap effect in minimizing chloride migration through silt. Iranian Journal of Science & Technology, 28 (B1), 107-118. Barone, F.S., Rowe, R.K., and Quigley, R.M. 1992. A laboratory estimation of diffusion and adsorption coefficients for several volatile organics in a natural clayey soil. Journal of Contaminant Hydrology, 10 (3), 225-250. Barone, F.S., Yanful, E.K., Quigley, R.M., and Rowe, R.K. 1989. Effect of multiple contaminant migration on diffusion and adsorption of some domestic waste contaminants in a natural clayey soil. Canadian Geotechnical Journal, 26 (2), 189-198. Benson, C.H. and Meers, S.R. 2009. Relative abundance of monovalent and divalent cations and the impact of desiccation on geosynthetic clay liners. Journal of Geotechnical and Geoenvironmental Engineering, 135 (3), 349-358. Bonaparte, R., Daniel, D.E., and Koerner, R.M. 2002. Assessment and Recommendations for Improving the Performance of Waste Containment Systems. EPA/600/R-02/099, U.S. Environmental Protection Agency, Cincinnati, Ohio, USA. Bouazza, A. and Rahman, F. 2004. Experimental and numerical study of oxygen diffusion through a partially hydrated needle-punched geosynthetic clay liner. Advances in Geosynthetic Clay Liner Technology, 2nd Symposium, ASTM STP 1456, R.E. Mackey and K. von Maubeuge, Eds., ASTM International, West Conshohoken, Pennsylvania, USA, 147-158. Bouazza, A. and Rahman, F. 2007. Oxygen diffusion through partially hydrated geosynthetic clay liners. Géotechnique, 57 (9), 767-772. Boving, T.B. and Grathwohl, P. 2001. Tracer diffusion coefficients in sedimentary rocks: Correlation to porosity and hydraulic conductivity. Journal of Contaminant Hydrology, 53 (1-2), 85-100. Bradshaw, S.L., Benson, C.H., and Scalia, J., IV 2013. Hydration and cation exchange during subsurface hydration and effect on hydraulic conductivity of geosynthetic clay liners. Journal of Geotechnical and Geoenvironmental Engineering, 139 (4), 526-538. Britton, J.P., Filz, G.M, and Little, J.C. 2005. The effect of variability in hydraulic conductivity on contaminant transport through soil-bentonite cutoff walls. Journal of Geotechnical and Geoenvironmental Engineering, 131 (8), 951-957. Çamur, M.Z. and Yazicigil, H. 2005. Laboratory determination of multicomponent effective diffusion coefficients for heavy metals in a compacted clay. Turkish Journal of Earth Sciences, 14, 91-103.

Cartwright, K. and Krapac, I.G. 1990. Construction and performance of a long-term earthen liner experiment. Waste Containment Systems: Construction, Regulation, and Performance, R. Bonaparte, Ed., ASCE, Reston, Virginia, USA, 135-155. Chapman, S.W. and Parker, B.L. 2005. Plume persistence due to aquitard back diffusion following dense nonaqueous phase liquid source removal or isolation. Water Resources Research, 41, W1241, doi:10.1029/2005WR004224. Charbeneau, R.J. and Daniel, D.E. 1993. Contaminant transport in unsaturated flow (Chapter 15). Handbook of Hydrology, D.R. Maidment, Ed., McGraw-Hill, New York, 15.1-15.54. Conca, J.L. and Wright, J. 1990. Diffusion coefficients in gravel under unsaturated conditions. Water Resources Research, 26 (5), 1055-1066. Cotten, T.E., Davis, M.M., and Shackelford, C.D. 1998. Effect of test duration and specimen length on diffusion testing of unconfined specimens. Geotechnical Testing Journal, 21 (2), 79-94. Crooks, V.E. and Quigley, R.M. 1984. Saline leachate migration through clay: A comparative laboratory and field investigation. Canadian Geotechnical Journal, 21 (2), 349-362. Cussler, E.L. 1997. Diffusion – Mass Transfer in Fluid Systems, 2nd ed., Cambridge University Press, Cambridge, United Kingdom. D'Appolonia, D.J. 1980. Soil-bentonite slurry trench cutoffs. Journal of Geotechnical Engineering Division, 106 (4), 399-417. De Soto, I.S., Ruiz, A.I., Ayora, C., Garcia, R., Regadio, M., and Cuevas, J. 2012. Diffusion of landfill leachate through compacted natural clays containing small amounts of carbonates and sulfates. Applied Geochemistry, 27 (6), 1202-1213. Demers, I., Bussiere, B., Mbonimpa, M., and Benzaazoua, M. 2009. Oxygen diffusion and consumption in lowsulphide tailings covers. Canadian Geotechnical Journal, 46 (4), 454-469. Devlin, J.F. and Parker, B.L. 1996. Optimum hydraulic conductivity to limit contaminant flux through cutoff walls. Ground Water, 34 (4), 719-726. Donahue, R.B., Barbour, S.L., and Headley, J.V. 1999. Diffusion and adsorption of benzene in Regina clay. Canadian Geotechnical Journal, 36 (3), 430-442. Edil, T.B. 2003. A review of aqueous-phase VOC transport in modern landfill liners. Waste Management, 23 (7), 561-571. Evangelou, V.P. and Zhang, Y.L. 1995. A review: Pyrite oxidation mechanisms and acid mine drainage prevention. Critical Reviews in Environmental Science and Technology, 2 (2), 141-199. Evans, J.C. 1991. Geotechnics of hazardous waste control systems. Foundation Engineering Handbook, 2nd Ed., H.Y. Fang, Ed., Van Nostrand Reinhold, NY, 750-777. Evans, J.C. 1993. Vertical cutoff walls (Chapter 17). Geotechnical Practice for Waste Disposal, D.E. Daniel, Ed., Chapman and Hall, London, 430-454. Evans, J.C. 1994. Hydraulic conductivity of vertical cutoff walls. Hydraulic Conductivity and Waste Contaminant Transport in Soil, ASTM STP 1142, D.E. Daniel and S.J. Trautwein, Eds., ASTM, West Conshohoken, Pennsylvania, USA, 79-94. Feenstra, S., Cherry, J. A., and Parker, B. L. 1996. Conceptual models for the behavior of dense nonaqueous phase liquids (DNAPLs) in the subsurface 19

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Katsumi, T., Benson, C.H., Foose, G.J., and Kamon, M. 2001. Performance-based design of landfill liners. Engineering Geology, 60 (1-4), 139-148. Kau, P.M.H., Binning, P.J., Hitchcock, P.W., and Smith, D.W. 1999. Experimental analysis of fluoride diffusion and sorption in clays. Journal of Contaminant Hydrology, 36 (1-2), 131-151. Khandelwahl, A., Rabideau, A.J., and Shen, P. 1998. Analysis of diffusion and sorption of organic solutes in soil-bentonite barrier materials. Environmental Science & Technology, 32 (9), 1333-1339. Klett, N.O. 2006. Leachate characterization and volatile organic compound (VOC) transport: A study of engineered landfills in Wisconsin. MS Thesis, University of Wisconsin-Madison, Madison, Wisconsin, USA. Krol, M.M. and Rowe, R.K. 2004. Diffusion of TCE through soil-bentonite slurry walls. Soil & Sediment Contamination, 13 (1), 81-101 Korf, E.P., Reginatto, C., Prietto, P.D.M., Thomé, A., and Consoli, N.C. 2011. Diffusive behavior of a compacted cemented soil as a containment barrier for industrial and mining waste. Geo-Frontiers 2011: Advances in Geotechnical Engineering, J. Han and D.E. Alzamora, Eds., ASCE, Reston, Virginia, USA, 926-936. Lake, C.B. and Rowe, R.K. 2000. Diffusion of sodium and chloride through geosynthetic clay liners. Geotextiles & Geomembranes, 18 (2-4), 103-131. Lake, C.B. and Rowe, R.K. 2005. Volatile organic compound diffusion and sorption coefficients for a needle-punched GCL. Geosynthetics International, 11 (4), 257-272. Lange, K., Rowe, R.K., and Jamieson, H. 2009. Diffusion of metals in geosynthetic clay liners. Geosynthetics International, 16 (1), 11-27. Lee, J. and Fox, P.J. 2009. Investigation of consolidationinduced solute transport. II. Experimental and numerical results. Journal of Geotechnical and Geoenvironmental Engineering, 135 (9), 1239-1253. Lee, J., Fox, P.J., and Lenhart, J.L. 2009. Investigation of consolidation-induced solute transport. I. Effect of consolidation on transport parameters. Journal of Geotechnical and Geoenvironmental Engineering, 135 (9), 1228-1238. Lever, D.A., Bradbury, M.H., and Hemingway, S.J. 1985. The effect of dead-end porosity on rock-matrix diffusion. Journal of Hydrology, 80 (1-2), 45-76. Lipson, D.S., Kueper, B.H., and Gefell, M. 2005. Matrix diffusion-derived plume attenuation in fractured bedrock. Ground Water, 43 (1), 30-39. Lorenzetti, R.J., Bartelt-Hunt, S.L., Burns, S.E., and Smith, J.A. 2005. Hydraulic conductivities and effective diffusion coefficients of geosynthetic clay liners with organobentonite amendments. Geotextiles & Geomembranes, 23 (5), 385-400. Mackay, D. M. and Cherry, J.A. 1989. Groundwater contamination: Pump-and-treat remediation. Environmental Science & Technology, 23 (6), 630-636. Malusis, M.A., Kang, J.-B., and Shackelford, C.D. 2013. Influence of membrane behavior on solute diffusion through GCLs. Coupled Phenomena in Environmental Geotechnics (CPEG 2013), July 1-3, 2013, Torino, Italy, CRC Press/Balkema, Leiden, The Netherlands, in press. Malusis, M.A., Maneval, J.E., Barben, E.J., Shackelford, C.D., and Daniels, E.R. 2010. Influence of adsorption on phenol transport through soil-bentonite vertical

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Neville, C.J. and Andrews, C.B. 2006. Containment criterion for contaminant isolation by cutoff walls. Ground Water, 44 (5), 682-686. Nguyen, T.-B., Lim, J., Choi, H., and Stark, T.D. 2011. Numerical modeling of diffusion for volatile organic compounds through composite landfill liner systems. KSCE Journal of Civil Engineering, 15 (6), 10331039. Nicholson, R.V., Gillham, R.W., Cherry, J.A., and Reardon, E.J. 1989. Reduction of acid generation in mine tailings through the use of moisture-retaining cover layers as oxygen barriers. Canadian Geotechnical Journal, 26 (1),1-8. Ogata, A. and Banks, R.B. 1961. A solution of the differential equation of longitudinal dispersion in porous media. U.S. Geological Survey Professional Paper 411-A. Olsen, H.W., Yearsley, E.N., and Nelson, K.R. 1990. Chemico-osmosis versus diffusion-osmosis. Transportation Research Record 1288, Geotechnical Engineering 1990, Transportation Research Board, National Research Council, National Academy Press, Washington, DC, USA, 15-22. Park, J.K. and Nibras, M. 1996. Mass flux of organic chemicals through polyethylene geomembranes. Water Environment Research. 65, 227-237. Park, J.K., Sakti, J.P, and Hoopes, J.A. 1996a. Transport of aqueous organic compounds in thermoplastic geomembranes. I. Mathematical model. Journal of Environmental Engineering, 122 (9), 800-806. Park, J.K., Sakti, J.P, and Hoopes, J.A. 1996b. Transport of aqueous organic compounds in thermoplastic geomembranes. II. Mass flux estimates and practical implications. Journal of Environmental Engineering, 122 (9), 807-813. Parker, B.L. and McWhorter, D.B. 1994. Diffusive disappearance of immiscible-phase organic liquids in fractured porous media: Finite matrix blocks and implications for remediation. Transport and Reactive Processes in Aquifers, T. Dracos and F. Stauffer, Eds., Balkema, Rotterdam, The Netherlands, 543-548. Parker, B.L., Cherry, J.A., and Gillham, R.W. 1996. The effects of molecular diffusion on DNAPL behavior in fractured porous media (Chapter 12). Dense Chlorinated Solvents and Other DNAPLs in Groundwater, J.F. Pankow and J.A. Cherry, Eds., Waterloo Press, Portland, Oregon, USA, 355-393. Parker, B.L., Gillham, R.W., and Cherry, J.A. 1994. Diffusive disappearance of immiscible-phase organic liquids in fractured geologic media. Ground Water, 32 (5), 805-820. Paumier, S., Touze-Foltz, N., Mazeas, L., and Guenne, A. 2011. Quantification of volatile organic compound diffusion for virgin geosynthetic clay liners and for a GCL after contact with a synthetic leachate. Journal of Geotechnical and Geoenvironmental Engineering, 137 (11), 1039-1046. Peters, G.P. and Smith, D.W., 2002. Solute transport through a deforming porous medium. International Journal of Numerical and Analytical Methods in Geomechanics, 26 (7), 683– 717. Polak, A., Nativ, R., and Wallach, R. 2002. Matrix diffusion in northern Negev fractured chalk and its correlation to porosity. Journal of Hydrology, 268 (14), 203-213. Quigley, R.M., Yanful, E.K., and Fernandez, F. 1987. Ion transfer by diffusion through clayey barriers.

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Management and Landfill Symposium, Italy, 245-254. Sangam, H.P. and Rowe, R.K. 2005. Effect of surface fluorination on diffusion through an HDPE geomembrane. Journal of Geotechnical and Geoenvironmental Engineering, 131 (6), 694-704. Sato, H. 1999. Matrix diffusion of simple cations, anions, and neutral species in fractured crystalline rocks. Nuclear Technology, 127 (2), 199-211. Sawatsky, N., Feng, Y., and Dudas, M.J. 1997. Diffusion of 1-naphthol and naphthalene through clay materials: Measurement of apparent exclusion of solute from the pore space. Journal of Contaminant Hydrology, 27 (12), 25-41. Scalia, J., IV and Benson, C.H. 2011. Hydraulic conductivity of geosynthetic clay liners exhumed from landfill final covers with composite barriers. Journal of Geotechnical and Geoenvironmental Engineering, 137 (1), 1-13. Seyedabbasi, M.A., Newell, C.J., Adamson, D.T., and Sale, T.C. 2012. Relative contribution of DNAPL dissolution and matrix diffusion to the long-term persistence of chlorinated solvent source zones. Journal of Contaminant Hydrology, 134-135, 69-81. Shackelford, C.D. 1988. Diffusion as a transport process in fine-grained barrier materials. Geotechnical News, 6 (2), 24-27. Shackelford, C.D. 1989. Diffusion of contaminants through waste containment barriers. Transportation Research Record No. 1219, Transportation Research Board, National Academy Press, Washington, DC, USA, 169-182. Shackelford, C. D. 1990. Transit-time design of earthen barriers. Engineering Geology, 29 (1), 79-94. Shackelford, C.D. 1991. Laboratory diffusion testing for waste disposal - A review. Journal of Contaminant Hydrology, 7 (3), 177-217. Shackelford, C.D. 1993. Contaminant transport (Chapter 3). Geotechnical Practice for Waste Disposal, D.E. Daniel, Ed. Chapman and Hall, London, 33-65. Shackelford, C.D. 1994. Critical concepts for column testing. Journal of Geotechnical Engineering, 120 (10), 1804-1828. Shackelford, C.D. 1995. Cumulative mass approach for column testing. Journal of Geotechnical Engineering, 121(10), 696-703. Shackelford, C.D. 1997. Modeling and analysis in environmental geotechnics: An overview of practical applications. 2nd International Congress on Environmental Geotechnics, IS-Osaka '96, M. Kamon, Ed., Balkema, Rotterdam, The Netherlands, Vol. 3, 1375-1404. Shackelford, C.D. 1999. Remediation of contaminated land: An overview. Proceedings, XI Pan-American Conference on Soil Mechanics and Geotechnical Engineering, Iguasu Falls, Brazil, Aug. 8-13, 1999, Vol. 4, 371-430. Shackelford, C.D. 2005. Environmental issues in International geotechnical engineering. 16th Conference on Soil Mechanics and Geotechnical Engineering, Millpress, Rotterdam, The Netherlands, Vol. 1, 95-122. Shackelford, C.D. 2011. Membrane behavior in geosynthetic clay liners. Geo-Frontiers 2011: Advances in Geotechnical Engineering, J. Han and D.E. Alzamora, Eds., ASCE, Reston, Virginia, USA, 1961-1970. Shackelford, C.D. 2012. Membrane behavior of engineered

clay barriers for geoenvironmental containment: State of the art. GeoCongress 2012-State of the Art and Practice in Geotechnical Engineering, R.D. Hryciw, A. Athanasopoulos-Zekkos, and N. Yesiller, Eds., ASCE, Reston, Virginia, USA, 3419-3428. Shackelford, C.D. 2013. Membrane behavior in engineered bentonite-based containment barriers: State of the art. Coupled Phenomena in Environmental Geotechnics (CPEG 2013), July 1-3, 2013, Torino, Italy, CRC Press/Balkema, Leiden, The Netherlands, in press. Shackelford, C.D, Benson, C.H., Katsumi, T., Edil, T.B., and Lin, L. 2000. Evaluating the hydraulic conductivity of GCLs permeated with non-standard liquids. Geotextiles & Geomembranes, 18 (2-4), 133161. Shackelford, C.D. and Daniel, D.E. 1991a. Diffusion in saturated soil. I: Background. Journal of Geotechnical Engineering, 117 (3), 467-484. Shackelford, C.D. and Daniel, D.E. 1991b. Diffusion in saturated soil: II. Results for compacted clay. Journal of Geotechnical Engineering, 117 (3), 485-506. Shackelford, C.D., Daniel, D.E., and Liljestrand, H.M. 1989. Diffusion of inorganic chemical species in compacted clay soil. Journal of Contaminant Hydrology, 4 (3), 441-473. Shackelford, C.D. and Jefferis, S.A. 2000. Geoenvironmental engineering for in situ remediation. International Conference on Geotechnical and Geoenvironmental Engineering (GeoEng2000), Melbourne, Australia, Nov. 19-24, Technomic Publ. Co., Inc., Lancaster, Pennsylvania, USA, Vol. 1, 121185. Shackelford, C.D., Malusis, M.A., and Olsen, H.W. 2003. Clay membrane behavior for geoenvironmental containment. Soil and Rock America Conference 2003, P.J. Culligan, H.H. Einstein, and A.J. Whittle, Eds., Verlag Glückauf GMBH, Essen, Germany, Vol. 1, 767-774. Shackelford, C.D. and Moore, S.M. 2013. Fickian diffusion of radionuclides for engineered containment barriers: Diffusion coefficients, porosities, and complicating issues. Engineering Geology, 152 (1), 133-147. Shackelford, C.D. and Nelson, J.D. 1996. Geoenvironmental design considerations for tailings dams. Proceedings, International Symposium on Seismic and Environmental Aspects of Dams Design: Earth, Concrete and Tailings Dams, Santiago, Chile, Oct. 14-18, Sociedad Chilena de Geotecnia, Vol. I. 131-187. Shackelford, C.D. and Redmond, P. 1995. Solute breakthrough curves for processed kaolin at low flow rates. Journal of Geotechnical Engineering, 121 (1), 17-32. Shackelford, C.D. and Rowe, R.K. 1998. Contaminant transport modeling. 3rd International Congress on Environmental Geotechnics, P. Seco e Pinto, Ed., Balkema, Rotterdam, The Netherlands, Vol. 3, 939956. Siegrist, R.L., Lowe, K.S., Murdoch, L.C., Case, T.L., and Pickering, D.A. 1999. In situ oxidation by fracture emplaced reactive solids. Journal of Environmental Engineering, 125 (5), 429-440. Sleep, B.E., Shackelford, C.D., and Parker, J.C. 2006. Modeling of fluid transport through barriers (Chapter 2). Barrier Systems for Environmental Contaminant Containment and Treatment, C.C. Chien, H.I. Inyang, 23

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and L.G. Everett, Eds., CRC Press, Taylor and Francis Group, LLC, Boca Raton, Florida, USA, 71-141. Smith, D.W., 2000. One-dimensional contaminant transport through a deforming porous medium: Theory and a solution for a quasi-steady-state problem. International Journal of Numerical and Analytical Methods in Geomechanics, 24 (8), 693– 722. Stormont, J.C., Morris, C.E., Finley, R.E. 1996. Capillary barriers for covering mine waste. 3rd International Conference on Tailings and Mine Waste '96, Balkema, Rotterdam, The Netherlands, 201-210. Struse, A. M., Siegrist, R. L., Dawson, H. E., and Urynowicz, M. A. 2002. Diffusive transport of permanganate during in situ oxidation. Journal of Environmental Engineering, 128 (4), 327-334. Tang, G., Alshawabkeh, A., and Sheahan, T.C. 2005. Experimental study of nonreactive solute transport in fine-grained soils under consolidation. Waste Containment and Remediation, A. Alshawabkeh, C.H. Benson, P.J. Culligan, J.C. Evans, B.A Gross, D. Narejo, K.R. Reddy, C.D. Shackelford, J.G. Zornberg, Eds., ASCE, Reston, Virginia, USA. Thoma, G.J., Reible, D.D., Valsaraj, K.T., and Thibodeaux, L.J. 1993. Efficiency of capping contaminated sediments in situ. 2. Mathematics of diffusion – adsorption in the capping layer. Environmental Science & Technology, 27 (12), 24122419. Toupiol, C., Willingham, T.W., Valocchi, A.J., Werth, C.J., Krapac, I.G., Stark, T.D., and Daniel, D.E. 2002. Long-term tritium transport through field-scale

compacted soil liner. Journal of Geotechnical and Geoenvironmental Engineering, 128 (8), 640-650. Touze-Foltz, N., Rosin-Paumier, S., Mazéas, L., and Guenne, A. 2011.Diffusion of volatile organic compounds through an HDPE geomembrane. GeoFrontiers 2011: Advances in Geotechnical Engineering, J. Han and D.E. Alzamora, Eds., ASCE, Reston, Virginia, USA, 1121-1130. Wang, X.Q., Thibodeaux, L.J., Valsaraj, K.T., and Reible, D.D. 1991. Efficiency of capping contaminated sediments in situ. 1. Laboratory scale experiments on diffusion–adsorption in the capping layer. Environmental Science & Technology, 25 (9), 15781584. Whitworth, T.M. and Ghazifard, A. 2009. Membrane effects in clay-lined inward gradient landfills. Applied Clay Science, 43 (2), 248-252. Willingham, T.W., Werth, C.J., Valocchi, A.J., Krapac, I.G., Toupiol, C., Stark, T.D., and Daniel, D.E. 2004. Evaluation of multidimensional transport through a field-scale compacted soil liner. Journal of Geotechnical and Geoenvironmental Engineering, 130 (9), 887-895. Xiao, S., Moresoli, C., Bolvenkamp, J., and De Kee, D. 1996. Sorption and permeation of organic environmental contaminants through PVC geomembranes. Journal of Applied Polymer Science, 63 (9), 1189-1197. Yanful, E.K. 1993. Oxygen diffusion through soil covers on sulphidic mine tailings. Journal of Geotechnical Engineering, 119 (8), 1207-1228.

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Schofield Lecture Centrifuge modelling: expecting the unexpected Conférence Schofield Modélisation physique en centrifugeuse: prévoir l'imprévisible Bolton M. D. Cambridge University

ABSTRACT: The unique advantage of physical modelling is that, unlike all forms of numerical simulation, it has the capacity to surprise its users with behaviour they would not have imagined. And the particular advantage of centrifuge testing is that observations are made on chosen soils, in a small format so that experiments can readily be repeated, and at magnitudes of stress and strain appropriate to field scale. However, it is the reasonable desire of centrifuge testers to represent their facilities as providing unambiguous predictions of field-scale performance through the application of accepted scaling laws, so as to recruit clients who will pay for such services. These diverse propositions create grounds for misunderstanding. Is centrifuge testing a cutting-edge research methodology capable of overthrowing conventional wisdom, or is it a well-understood tool capable of unambiguously recreating field-scale behaviour? This question sets the theme for the paper. In attempting to answer it, a variety of geotechnical modelling issues will be explored, including cyclic shearing and excess pore pressures, localisation and cracking, creep and strain-rate effects, and the possible influence of grain size and soil structure. In doing so, the key concept will be that of a behavioural mechanism. Weaker associations that may be made between a model, its prototype and a real field-scale structure will then be scrutinised.

RÉSUMÉ : Le principal avantage de la modélisation physique est que, contrairement à la modélisation numérique, elle peut surprendre l’utilisateur avec des résultats qu’il n’aurait pu imaginer. Pour la modélisation physique en centrifugeuse, cet avantage est augmenté par le fait que les sols utilisés ont été choisis par l’utilisateur, que les expériences sont réalisées à petite échelle et peuvent être facilement répétées et que les niveaux de contraintes sont identiques à ceux rencontrés à échelle réelle. Cependant, il est légitime pour chaque utilisateur d’espérer que les résultats de ses observations expérimentales puissent être extrapolés sans ambiguïté aux structures réelles qu’il cherche à modeler, grâce notamment à l’utilisation de lois de similitude parfaitement établies, afin de pouvoir attirer d’éventuels clients et de financer ses recherches. Ces différentes observations peuvent mener à de profondes incompréhensions. La modélisation physique en centrifugeuse est-elle un outil de recherche avancé capable de bouleverser notre compréhension des phénomènes géotechniques, ou est-ce un instrument parfaitement maîtrisé, capable de modéliser sans ambiguïté le comportement des structures réelles ? Cette question est le thème principal de cet article. En tentant d’y répondre, un vaste de champs de problèmes sera abordé, incluant notamment les problèmes associés au cisaillement cyclique, à la génération de pressions interstitielles, aux déformations différées, aux effets de vitesse de cisaillement, et à la possible influence de la taille des grains sur l’interaction sol structure. Ce faisant, le concept clef de mécanisme de comportement sera énoncé. D’autres éléments permettant d’associer les modèles, les prototypes et les structures réelles seront également étudiés.

KEYWORDS: centrifuge testing, models, scaling laws, mechanisms.

The written contribution was not received at time of editing the Proceedings. La contribution écrite n’a pas été fournie avant l’édition des Actes.

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Enjeux géotechniques pour la construction du métro automatique « Grand Paris Express » Geotechnical issues for « Grand Paris Express » automatic metro Fluteaux V. Société du Grand Paris

RÉSUMÉ : La Société du Grand Paris, Établissement Public d’État, a pour mission de concevoir et de réaliser le réseau de transport public « Grand Paris Express » qui est au cœur du projet d’aménagement du territoire : « Nouveau Grand Paris ». Ce projet comporte 4 nouvelles lignes, 2 extensions : il se développe sur 200 km et comprend 72 gares, principalement en souterrain. Dans un premier temps, il est détaillé le processus d’utilisation des données et la mise en place de regards partagés entre les différents acteurs (maître d’ouvrage, assistants à maîtrise d’ouvrage, maîtres d’œuvre) sur ces dernières afin de maîtriser les risques techniques liés notamment à la géotechnique et à la reconnaissance du bâti. Dans un second temps, sont listés les enjeux liés à la géotechnique pour le projet du Grand Paris Express et comment ces enjeux ont été pris en compte dès les premières phases d’études amont, et ont, de ce fait, orienté, les investigations géotechniques qui sont résumées en quelques chiffres. Ces enjeux sont les carrières anthropiques, les nappes d’eau, la dissolution du gypse et le retrait gonflement des argiles. ABSTRACT: « La Société du Grand Paris », a Public company, was created for the conception and construction of « Grand Paris Express » transport network, the major link of « Nouveau Grand Paris » urban development. This project is composed of 4 new lines and 2 line extensions, with 72 stations: its total length is 200 km, mainly underground level. In a first time, the method used for collecting geotechnical and existing building data is exposed, also the corresponding risk management approach of the owner, his advisors and the designers. In a second time, the main geotechnical issues are identified and mitigated at the first conception phase with pertinent geotechnical investigations: existing underground quarries, water table layers, gypsum dissolution and clay swelling. MOTS-CLÉS : développement urbain, souterrain, reconnaissances géotechniques, maîtrise des risques KEYWORDS: Urban development, tunnelling, geotechnical investigation, risk management

1. INTRODUCTION

Il comporte :

Le Nouveau Grand Paris - Le projet du Grand Paris Express Le projet du Nouveau Grand Paris c’est d’abord un projet d’aménagement du territoire. L’État a affirmé par là sa volonté de valoriser les territoires de l’Île-de-France et d’améliorer la capacité de logement. Ce projet répond à trois enjeux majeurs : améliorer la vie quotidienne des Franciliens (enjeu de qualité de vie), favoriser le désenclavement des territoires (enjeu de solidarité) et leur développement économique (enjeu d’attractivité et d’emploi). Le Grand Paris Express contribue à structurer ce grand projet d’aménagement, de par l’amélioration à court terme du service offert aux voyageurs, la modernisation et l’extension du réseau existant mais également par la création d’un réseau de transport public automatique qui concerne l’ensemble de l’Île-de-France. Il vise à desservir de nouveaux territoires et à apporter enfin une réponse satisfaisante aux très nombreux voyageurs qui vont quotidiennement de banlieue à banlieue et qui sont à ce jour obligés de transiter par Paris. Il a aussi vocation à pouvoir relier les différents aéroports à l’ensemble des activités de la région IDF.

•

une rocade de grande capacité, la ligne 15, désaturant la zone dense ;

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des transports automatiques à capacité adaptée pour la desserte des territoires en développement : ligne 16, ligne 17, ligne 18 ;

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des prolongations de lignes de métro existants: ligne 14 au nord et au sud; ligne 11 vers l’est, de Mairie des Lilas à Noisy-Champs via Rosny-Bois-Perrier.

Ces trois lignes représentent 166 km de métro et 57 gares ; avec la ligne orange (réseau complémentaire), cela représente 200 km et 72 gares. L’ensemble, le Grand Paris Express, constitue un réseau très maillé avec les infrastructures de transport existantes (métro, RER, Transilien) afin de fluidifier les échanges à l’échelle de l’Île-de-France. Le schéma d’ensemble de ce projet a fait l’objet d’un décret en Conseil d’État en août 2011.

Le projet du Grand Paris Express a fait l’objet en janvier 2011 d’un protocole d’accord entre l’État et la région Île-deFrance, il représente une synthèse des projets portés antérieurement par les deux autorités.

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Une procédure d’enquête publique sur un premier tronçon de la ligne rouge (future ligne 15, entre Noisy-Champs et Pont de Sèvres) a été lancée. Les premières mises en service sont attendues à l’horizon 2020 sur ce tronçon.

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La Société du Grand Paris

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La Société du Grand Paris (SGP), Établissement Public de l’État, a été créée par une loi de 2010. Elle est Maître d’Ouvrage du réseau de transport public du Grand Paris Express. Sa mission est de concevoir et de réaliser ce nouveau réseau. La SGP a également une capacité d’aménager ou de construire autour des gares. La Société du Grand Paris est constituée d’équipes pluridisciplinaires, avec notamment une direction du programme qui regroupe des équipes de projet et des unités métiers qui assurent une approche transversale. Avec plusieurs dizaines de gares concernées par le projet, il est en effet nécessaire de définir des règles communes et transversales de conception. De la même manière, le projet étant essentiellement souterrain et parfois assez profond, une équipe de spécialistes de travaux souterrains intervient en conseil auprès des équipes projet, elle définit et gère notamment les compagnes de reconnaissance du sous-sol et du bâti à l’échelle du nouveau réseau. Le présent article vise à détailler la mission du Maître d’Ouvrage durant les phases amont d’acquisition des données, et comment ces données vont être utilisées et par quel acteur par la suite. La géologie du Bassin Parisien est plutôt bien connue et dès les premières études du projet, le MOA disposait d’une base documentaire sérieuse et assez étoffée sur les couches géologiques intéressées par le projet (INFOTERRE). Il est bien évident que cela n’est pas suffisant et que des campagnes de reconnaissances ont dû être planifiées et dont certaines sont en cours de réalisation. S’ils servent d’abord à définir les paramètres servant aux calculs de dimensionnement des ouvrages, ces résultats doivent permettre aussi d’apprécier le comportement des terrains lors de l’exécution des travaux et les conséquences sur l’environnement, et en tout en premier lieu sur le bâti. La typologie et l’état de ce bâti sont donc évidemment déterminants pour juger des conséquences des travaux. Des enquêtes détaillées du bâti seront donc entreprises. En attendant, des investigations de terrain et documentaires ponctuelles, a minima qualitatives, ont été effectuées toutes les fois que la faisabilité de la réalisation des ouvrages du projet était en jeu. Une bonne part des reconnaissances réalisées à ce jour a permis d’alimenter les études de conception « amont », plus particulièrement les études préliminaires qui se sont déroulées tout au long de l’année 2012.

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Les effets négatifs sur les ouvrages existants liés à la réalisation de nouveaux ouvrages enterrés sont bien sûr très variables selon l’ouvrage concerné et son état. Ces effets vont de l’atteinte au fonctionnement normal de celui-ci jusqu’à sa dégradation voire sa ruine. Cette plus ou moins grande sensibilité du bâti, des infrastructures et des réseaux existants aux travaux de réalisation du projet est également fonction de la nature et de la qualité des terrains rencontrés et des éventuelles contraintes que sont la nature du sous-sol, la présence de vides dans le sol ou de décompressions préexistantes etc… ainsi que de la profondeur du tunnel. Il est donc essentiel dès les premières phases de faire un recensement de qualité, mission qui incombe à la Maîtrise d’Ouvrage. 2.1.2

Dès les premières phases de conception du projet, la Société du Grand Paris s’est donc attachée à recenser l’ensemble des contraintes susceptibles d’interférer avec le projet : recherche des réseaux enterrés ou infrastructures, établissement d’un diagnostic des zones traversées tant du point de vue du sous-sol que du point de vue de l’état du bâti (y compris réseaux enterrés et infrastructures).

2.1.1

Organisation des études sur le bâti, les réseaux enterrés et les infrastructures

Dès les phases amont (études de faisabilité et études préliminaires) conduites par le Maître d’Ouvrage, les réseaux structurants (non déviables) ont fait l’objet d’un recensement bibliographique en partenariat avec les différents concessionnaires concernés : RATP, SNCF, égouts, transports d’énergie, etc. De même, concernant le bâti, ont été recensés les bâtiments susceptibles d’interférer avec le projet (immeubles de grande hauteur, fondations profondes…). Ainsi, les premiers tracés réalisés ont tenu compte, tant en plan qu’en profil, de ces contraintes et n’interfèrent pas avec ces grands réseaux ou obstacles. Les études de maîtrise d’œuvre à venir vont permettre d’affiner les connaissances sur ce bâti, les objectifs sont multiples :  confirmer et/ou compléter le recensement des études préliminaires des grands réseaux non déviables, afin d’établir un tracé prenant en compte l’ensemble de ces contraintes ;  établir la méthodologie des travaux de confortement à entreprendre en cas de proximité de ces grands réseaux ;  identifier, concevoir et initialiser les déviations de réseaux en amont des travaux de génie civil pour les réseaux déviables ;  caractériser le bâti dans la zone d’interférence du projet, dans le but de déterminer sa sensibilité. Pour atteindre ces objectifs, deux démarches doivent être menées : Le recensement systématique des réseaux présents sur le tracé : Ce recensement porte sur l’exhaustivité des réseaux (déviables et non déviables). Il permettra notamment de caractériser les réseaux tant géométriquement (localisation en plan et en profondeur) que qualitativement (nature, état de conservation et fonctionnement des réseaux). Une enquête sur le bâti et les infrastructures couplée à une étude de sensibilité : La Société du Grand Paris va s’adjoindre les conseils d’un Assistant à Maîtrise d’Ouvrage en expertise du bâtiment. Cette

2. LES GRANDS AXES DE CONCEPTION

2.1

Bâti - tous les types de bâtiments sont présents. Leur tolérance aux déformations du sol qui pourraient être provoquées par l’exécution d’un projet de métro souterrain dépend du type de construction et du type de fondations du bâtiment. Réseaux enterrés - seuls les réseaux de taille importante, ne pouvant être déviés, représentent un véritable enjeu pour le projet, à savoir notamment : les canalisations d’assainissement, transports énergie (gaz, pétrole) et les canalisations de chauffage urbain. Infrastructures - sont notamment concernés les ouvrages d’art et les infrastructures ferroviaires, routières.

Le bâti, les réseaux enterrés et les infrastructures Types d’ouvrages rencontrés à proximité du projet et enjeux liés à leur présence

Le projet a potentiellement une influence sur différents types d’ouvrages :

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enquête et cette étude de sensibilité seront réalisées dans la zone d’influence géotechnique du projet. Elle aura un double objectif : reconnaître le bâti au sens large tant d’un point de vue géométrique que structurel (niveau des fondations, système de poutraison, etc.) ; mais également déterminer sa vulnérabilité (tolérance aux déformations du sol) aux travaux envisagés. Ces données d’entrée seront ensuite fournies au Maître d’Œuvre pour prise en compte dans la conception du réseau. Sur la base de l’analyse de ces données, il conviendra d’adapter le dimensionnement des ouvrages du métro et/ou les méthodes constructives de manière à respecter les tolérances des ouvrages existants afin de réduire voire supprimer le risque. Le but de cette organisation est d’avoir un regard partagé sur l’interprétation des données entre l’assistant à Maîtrise d’Ouvrage bâti et le maître d’œuvre afin de concevoir un projet adapté au contexte de sensibilité du bâti présent dans la zone d’influence géotechnique. Dans cette organisation l’assistant à Maîtrise d’Ouvrage Géotechnique a bien sûr un rôle essentiel à jouer (cf.2.2.2). De plus, la Société du Grand Paris dès les phases d’études de Maîtrise d’Œuvre va mettre en place un Comité de maîtrise des risques qui sera constitué d’experts indépendants. Ce comité sera consulté sur les grandes orientations techniques du projet, mais également sur les points sensibles. A travers cette organisation tournée vers l’expertise des sujets sensibles, dont fait notamment partie la caractérisation du bâti pour la détermination des méthodes constructives, la Société du Grand Paris entend maîtriser la qualité technique, les risques, les coûts et les délais. 2.1.3

Dispositions mises en place en phase travaux

En complément, afin de vérifier que les mesures retenues lors des différentes études réalisées permettent bien de supprimer les risques d’impact sur le bâti, une auscultation de celui-ci sera mise en place le long du tracé dans les zones sensibles : cette auscultation sera mise en place en amont des travaux, afin de mesurer la respiration naturelle des ouvrages liée notamment aux variations thermiques ; en phase chantier, une surveillance de l’existant en temps réel sera mise en œuvre, le but étant de comparer les déformations estimées aux déformations observées afin de pouvoir adapter les méthodes constructives immédiatement en cas de déplacement jugé anormal. Comme dans la phase de conception, cette auscultation fera l’objet d’un double regard entre l’assistant à Maîtrise d’Ouvrage en bâti et le maître d’œuvre, ainsi que d’une expertise éventuelle du Comité de Maîtrise des Risques. 2.2 2.2.1

La géologie, l’hydrogéologie et la géotechnique Le but des investigations géotechniques entreprises

Un projet de transport en souterrain est, par essence, en forte interaction avec le sous-sol. De ce fait, afin de réaliser des études de qualité, la connaissance parfaite du sous-sol au sens large est nécessaire, les investigations géotechniques entreprises dès la phase d’études préliminaires ont classiquement pour objectifs : D’établir le modèle géologique du projet : coupe linéaire par corrélation entre les points de sondages. D’établir un modèle hydrogéologique. Les investigations doivent permettre de caractériser le ou les aquifères en présence, tant d’un point de vue piézométrique que d’un point de vue perméabilité.

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De caractériser les couches rencontrées tant d’un point de vue mécanique (dimensionnement les ouvrages de génie civil) qu’environnemental (détermination de la destination d’évacuation des déblais). Le but final est de localiser et caractériser des zones dites « homogènes» afin d’adapter les méthodes constructives à chacune d’entre elles. Des zones singulières peuvent également être identifiées (exemple zone de dissolution de gypse), qui feront l’objet de reconnaissances spécifiques au regard de la singularité rencontrée, permettant ainsi de mettre en place les méthodes constructives et les confortements adaptés. 2.2.2

Organisation des études géotechniques

Les études géotechniques sont régies par la norme NF P 94500 relative aux missions géotechniques. Ces missions sont à mettre en regard des phases d’études de conception définies par la loi relative à la Maîtrise d’Ouvrage Publique « loi MOP », cf. le tableau ci-après qui récapitule les caractéristiques de chacune des phases :

Phases d’études Loi MOP

Phases d’études géotechniques (NF P 94500)

Etudes de Faisabilité Etudes Préliminaires

Nature de la donnée

Dossier à remettre

Mission G11 Phase 1

Bibliographique

Mission G11 Phase 2

Reconnaissances sur site

Premier modèle géologique, hydrogéologique Première identification des risques.

Production du dossier d’enquête publique Phase d’AvantMission G12 Reconnaissances Projet sur site

Phase Projet

Mission G2

Reconnaissances sur site

Identification des aléas majeurs et principes généraux pour en limiter les conséquences Identifications des aléas importants et dispositions pour en réduire les conséquences

C’est toujours au travers d’une organisation rigoureuse, permettant divers niveaux d’expertises, que la Société du Grand Paris compte maitriser les risques techniques (dans un projet de travaux en souterrain, ils sont essentiellement liés au sol), les coûts et les délais. Pour se faire, la Société du Grand Paris s’est adjoint les conseils d’un assistant à maîtrise d’ouvrage en géotechnique, qui a plusieurs missions : Définir et superviser les investigations géotechniques, Interpréter et établir pour le compte de la Société du Grand Paris les missions G11, G12 et G2, Accompagner la Maîtrise d’Ouvrage dans ses discussions avec le maître d’œuvre. Les résultats factuels de ces investigations géotechniques sont transmis au maître d’œuvre pour une analyse et une interprétation qui lui sont propres, ce qui double la réalisation des missions G12 et G2. Le but de cette organisation est d’avoir un regard partagé sur l’interprétation des données de sols entre les spécialistes du maître d’œuvre et l’assistant à maîtrise d’ouvrage Géotechnique, afin de concevoir un projet adapté au contexte géologique, hydrogéologique et géotechnique par une adéquation des méthodes constructives retenues. De plus, le Comité de maîtrise des risques, sera consulté dans tous les grands choix techniques qui sont liés à la géotechnique et aux méthodes constructives.

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2.2.3

o piliers à bras (poteaux montés pour soutenir le ciel de carrière), o par hagues et bourrages : réalisation de murs de pierres sèches (hagues) retenant les déchets non utilisés de l’exploitation de la carrière (bourrages). Ces confortements quels qu’ils soient, et quelle qu’en soit la qualité de réalisation, laissent subsister des vides.

Dispositions mises en place en phase travaux

La norme NF P 94-500 régit également la phase de réalisation, et impose la réalisation de deux missions G3 et G4 : La mission G3 : mission d’études et de suivi géotechnique des travaux portée par l’entreprise qui réalise les travaux, La mission G4 : mission de supervision géotechnique d’exécution portée par le Maître d’ouvrage, et déléguée à son Maître d’œuvre. La SGP a prévu d’être accompagnée de spécialistes dans le domaine de la géotechnique via des missions de conseil, pour maintenir la vision partagée sur les données géotechniques et les méthodes constructives qu’elle a initié dès les phases amont du projet. De plus, le Comité de maîtrise des risques, sera consulté dans tous les grands choix techniques qui sont liés à la géotechnique et les méthodes constructives.

3.1.2

Dès la fin de leur exploitation, ces carrières ont été le siège de mouvements verticaux pouvant entraîner des cloches de décompressions dans les terrains sus-jacents, voire dans le cas extrême la formation de fontis. Les carrières à ciel ouvert sont souvent remblayées par des matériaux de mauvaise qualité et présentent donc l’inconvénient de constituer des terrains médiocres, sous-consolidés pour la réalisation d’un projet de type métro souterrain. D’une part, la traversée de ces terrains meubles peut perturber le bon avancement du tunnelier. D’autre part, le passage du tunnelier dans des terrains sous-consolidés peut engendrer des tassements en surface difficilement compatibles avec le bâti de surface. L’enjeu est donc d’autant plus fort en présence de bâtis denses ou d’ouvrages particuliers. Les anciennes carrières souterraines constituent des ouvrages fragiles. Le passage du tunnelier ou la réalisation de travaux à proximité de ces dernières, et donc la modification du milieu en termes de contraintes dans le sol notamment, peut engendrer la remise en cause de l’équilibre précaire des carrières. Ainsi, la réalisation d’un projet de type métro souterrain à proximité d’anciennes carrières souterraines est susceptible de créer des désordres sur ces dernières, engendrant des décompressions dans le sol et donc des tassements pouvant remonter en surface et impacter le bâti, situé à l’aplomb des zones concernées.

3. LES PRINCIPAUX ENJEUX GÉOLOGIQUES, HYDROGÉOLOGIQUES ET GÉOTECHNIQUES Le projet de réalisation du réseau Grand Paris Express s’insère majoritairement en souterrain ; il traverse des nappes d’eaux souterraines et des couches géologiques aux caractéristiques très diverses. Des études préliminaires et des sondages entrepris dans ce cadre, il ressort que les principaux enjeux géologiques, hydrogéologiques et géotechniques du projet sont les suivants : - prendre en compte les cavités d’origine anthropique (résultant des activités humaines) que sont les anciennes carrières. Le projet passe sous plusieurs anciennes carrières souterraines et à ciel ouvert. Cet enjeu représente l’une des priorités auxquelles les études de conception se sont attachées à répondre (forte concentration de carrières dans le périmètre du projet sur la partie sud principalement). - éviter au maximum tout impact sur les nappes d’eaux souterraines, que ce soit en termes de pollution des eaux, de modification du niveau des nappes ou de modification de la circulation des eaux. - identifier et prendre en compte les zones marquées par la présence de sols évolutifs (horizons contenant du gypse), dans lesquels des phénomènes de dissolution peuvent avoir lieu. Cet enjeu est principalement localisé au nord et nord-est du réseau. - prendre en compte la présence d’argiles et par conséquent un phénomène éventuel de retrait ou de gonflement des argiles. Cet enjeu se révèle toutefois assez mineur et très localisé à l’échelle du projet.

Il est important de souligner que le risque lié aux carrières sur un chantier de type métro provient essentiellement de carrières qui n’auraient pas été identifiées préalablement au chantier. Il est donc primordial de connaître parfaitement leur localisation, leur étendue et leur état. 3.1.3

Prise en compte des enjeux dès la conception

Le meilleur moyen de supprimer les risques liés à la présence de carrières est de les contourner ou de s’en éloigner au maximum. De façon générale, le tracé en plan du projet cherche autant que possible à éviter la traversée de zones de carrières en s’en éloignant au maximum (démarche identique pour les carrières souterraines et à ciel ouvert), lorsque cela était compatible avec les objectifs de desserte du projet. Lorsque la zone de carrière n’a pu être évitée : - Pour les carrières à ciel ouvert remblayées, le profil en long du tunnel est adapté afin que l’épaisseur de terrain audessus de la voûte du tunnel soit suffisante pour que les tassements soient non significatifs pour le bâti sus-jacent. En cas de remblaiement très médiocre de la carrière à ciel ouvert, des traitements de terrains peuvent être mis en place. Les études à venir permettront d’identifier et de caractériser ces remblais afin d’adapter au mieux le passage du tunnelier dans ces zones. - Pour les carrières souterraines, le profil en long du tunnel est ajusté afin de le faire passer sur la majorité des zones concernées en dessous de ces dernières. En effet, le passage au travers d’une carrière souterraine est délicat, du fait de son équilibre précaire, du manque d’homogénéité des terrains traversés et de leur mauvaise qualité, et doit donc se cantonner à des linéaires très faibles nécessitant de ce fait des confortements préalables lourds. D’autre part, pour une grande partie du tracé

3.1 Enjeux particuliers liés à la présence d’anciennes carrières 3.1.1

Enjeux d’un projet de type métro souterrain liés à la présence de carrières

Types de carrières rencontrés

La présence de carrières fait l’objet de Plan de Prévention des Risques à l’échelle de l’Île-de-France. Le secteur sud de Paris a été largement exploité pour la construction de Paris jusqu’au XIXe siècle, essentiellement à partir de carrières souterraines. On rencontre actuellement d’anciennes carrières qui ont servi à l’extraction de matières premières variées (le calcaire grossier en pierre à bâtir ; le gypse pour le plâtre ; les marnes, craies pour le ciment et la chaux, et les sables pour l’industrie…). Les carrières à proximité du projet se divisent en deux principales familles : • les carrières à ciel ouvert, remblayées après leur exploitation par du tout-venant, • les carrières souterraines avec plusieurs techniques de confortement possibles :

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l’épaisseur de terrain n’est pas suffisante pour faire passer le tunnel au-dessus de la zone de carrières souterraines La distance entre la voûte du tunnel et le plancher des carrières est ajustée selon la nature du terrain, afin de conserver une épaisseur suffisante de « bon » terrain au-dessus du tunnel. Les hypothèses prises en compte en études préliminaires seront à confirmer dans les études à venir qui permettront d’affiner l’identification et la caractérisation des terrains situés en dessous des carrières afin de définir la bonne distance à retenir entre la voûte et la base de carrière. Si la carrière s’avérait en trop mauvais état ou si la distance entre la voûte et la base de la carrière était trop faible, des traitements devront être réalisés, afin que la stabilité d’ensemble du massif soit préservée. Les études et les investigations à venir permettront d’identifier ces zones. 3.1.4

Mesures mises en œuvre

Les mesures à mettre en œuvre pour supprimer le risque de désordre sur les bâtis et les ouvrages souterrains dans la zone d’influence du projet et des carrières sont les suivantes: - En phase études : Investigations des anciennes carrières avant le chantier (bibliographie, visites, inspections, sondages, essais, mesures in situ) afin de reconnaître leurs limites, leurs épaisseurs, la nature des remblais de comblement et de définir l’état de la carrière. Ces investigations ont pour objet de caractériser le massif et ainsi définir les zones et les volumes à traiter, ainsi que le type de traitement à mettre en place. - En phase travaux, pour les zones où les études ont montré la nécessité d’un traitement de carrières : des injections ou comblements des carrières (à ciel ouvert ou souterraines) nécessitant un confortement pourront être réalisées. Plusieurs techniques sont possibles ; le traitement retenu dépendra de différents paramètres dont la distance entre le plancher de la carrière et la voûte du tunnel, le mode de stabilisation préexistant de la carrière, l’état de la carrière, la densité du bâti en surface, la nature des terrains, etc... Deux grandes techniques existent : • injection depuis la surface grâce à des forages afin de combler les carrières avant la réalisation du tunnel ; • comblement à pied d’œuvre : réalisation du comblement depuis les galeries des carrières, mise en place de murs masques et remplissage par mortier à l’arrière. 3.1.5

Méthodes de suivi des effets des mesures

Afin de vérifier l’efficacité du renforcement des carrières, des sondages de contrôle des traitements des carrières seront réalisés. Par ailleurs, une méthode observationnelle sera mise en place, comme décrit au chapitre 2.1 3.2 Nappes et circulations souterraines Le projet Grand Paris Express de par sa profondeur va se situer majoritairement sous nappe. De ce fait, sa réalisation est susceptible de générer différents phénomènes : - modification du niveau de la (des) nappe(s), - modification des écoulements : « effet barrage », - pollution d’une nappe par mise en communication, du fait de la réalisation de l’infrastructure, avec une nappe polluée. Le projet Grand Paris Express va traverser différents aquifères ayant leur propre système hydrogéologique : sens d’écoulement, puissance de l’aquifère, perméabilité. Chaque aquifère ne répondra pas de la même façon à la réalisation de ce projet.

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Impacts de la modification du régime d’écoulement des eaux souterraines

Une modification du régime d’écoulement des eaux souterraines, quelle qu’en soit l’origine, pourrait avoir un certain nombre d’incidences potentielles: - Tassements : l’abaissement du niveau de la nappe peut générer des déformations de terrain en surface. - Ennoiement des structures enterrées : la hausse du niveau de la nappe peut provoquer des ennoiements dans les sous-sols de bâti existant. - Une modification des écoulements et de la teneur en eau des sols peut favoriser différents phénomènes naturels tels que la dissolution du gypse (cf. titre 3.3) ou le retrait/gonflement des argiles (cf. titre 3.4). - Modification des usages anthropiques : l’abaissement du niveau de la nappe risque notamment de dénoyer les pompes existantes (captage d’eau potable, géothermie, etc.). 3.2.2

Prise en compte des enjeux dès la conception

Modification du niveau de la/des nappe(s) Le tunnel n’impacte pas le niveau des eaux souterraines. En effet, la technique du tunnelier permet d’éviter tout rabattement de nappe en créant une paroi étanche à l’avancement de l’excavation. Pour les gares et tranchées couvertes, la méthode constructive « enceinte étanche » en parois moulées est majoritairement retenue, ce qui limite les venues d’eau horizontales. En fonction des conditions géologiques (zones à dissolutions de gypse potentielles par exemple), de la sensibilité de la zone en termes environnementales (proximité de zones abritant des espèces protégées ne permettant pas une modification du niveau piézométrique, même temporaire), il peut également s’avérer nécessaire de limiter au maximum les venues d’eau par le fond de fouille. Pour cela un bouchon injecté en sous face du radier sera réalisé. Dans le cas de terrain imperméable en fond de fouille ou de bouchon injecté, l’eau extraite en phase chantier se limite au volume d’eau contenu dans la gare, ainsi que des venues d’eau résiduelles, l’évacuation de ces dernières est toutefois non significative au regard du niveau de la nappe baignant l’ouvrage. Certaines gares peuvent être réalisées en technique mixte associant la réalisation d’un puits en parois moulées présentant les mêmes caractéristiques que présentées ci-avant, et le reste réalisée en méthode traditionnelle. Cette technique est utilisée lorsque des contraintes de surface ne permettent pas de réaliser la gare uniquement depuis la surface. La partie puits est similaire au cas d’une enceinte étanche. Pour la partie traditionnelle des pompages en phase chantier s’avèreront nécessaires ; quand le contexte géologique ou environnemental ne permettra pas de rabattre la nappe de façon importante, un traitement d’étanchéité préventive des terrains sera mis en place par injections ou jet-grouting, permettant ainsi de limiter au maximum l’impact des travaux sur le niveau piézométrique de la nappe en présence. Effet barrage L’effet barrage induit par la réalisation du tunnel est fonction du sens d’écoulement de la nappe dans laquelle s’insère ce dernier. Cet effet se manifeste par l’abaissement du niveau piézométrique en aval du tunnel et une augmentation du niveau piézométrique en amont de ce dernier. Certaines zones du tracé ont d’ores et déjà été identifiées comme susceptibles d’être le siège d’un tel phénomène (Pont de Sèvres, le Nord de Paris). Certaines gares ou tranchée couverte peuvent induire un effet barrage non négligeable, qui nécessite la mise en œuvre de mesures spécifiques. En premier lieu, les études à venir permettront de quantifier cet effet et d’estimer le réel impact des ouvrages du métro souterrain sur le niveau des nappes afin de

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prendre d’éventuelles dispositions pour limiter ce phénomène. Il existe différents dispositifs permettant de lutter contre cet effet de barrage parmi lesquels : recepage des têtes de parois moulées, tranchées drainantes, siphons, etc. Mise en communication des nappes La mise sous pression de la chambre d’abattage en tête de tunnelier et l’injection du vide annulaire permettent d’éviter les arrivées d’eau à l’intérieur de l’ouvrage. La technologie utilisée de foration au tunnelier limite donc le risque de communication entre nappes. La technique des parois moulées utilisée pour la majorité des gares permet de limiter grandement les échanges entre les nappes grâce à la mise sous pression de la fouille par la boue bentonitique au moment du creusement. Pour les parties réalisées en méthode souterraine traditionnelle, les pré-injections d’étanchement permettent de limiter grandement les échanges entre nappes. Les vides qui potentiellement pourraient subsister suite à la mise en place des structures définitives pourront faire l’objet d’injections de remplissage afin de minimiser les échanges entre nappes, si cela s’avérait nécessaire. 3.2.3

3.3.2

Dans un premier temps, afin de maîtriser les effets de la présence de gypse, les mesures suivantes seront mises en œuvre dans les zones concernées : - reconnaissances adaptées avant le chantier, en phase études (bibliographie, sondages, mesures géophysiques, analyses hydrogéologiques…) : le but est de caractériser le massif et de repérer d’éventuelles décompressions ou vides ; - si des anomalies étaient repérées, il pourrait être nécessaire de traiter les vides par injection, selon leur taille et la sensibilité de la zone d’influence du creusement (présence de bâti, d’ouvrage sensible, etc.). Dans un deuxième temps, afin d’éviter les phénomènes de dissolution du gypse, la conception du projet Grand Paris Express prévoit de limiter au maximum l’impact du projet sur le régime d’écoulement des nappes dans les zones susceptibles de développer ce phénomène. La conception et les mesures adoptées pour limiter cet impact sont donc celles décrites au chapitre 3.2.2 récapitulant les méthodes constructives à mettre en œuvre pour ne pas perturber le niveau piézométrique et les sens d’écoulement des nappes.

Méthodes de suivi des effets des mesures 3.3.2

Un suivi des mesures mises en œuvre pour supprimer les impacts du projet sur le régime des nappes et eaux souterraines est mis en place dès la phase étude. En particulier : - un « état zéro » est établi pour les différents paramètres (niveau piézométrique, débits, température, pH, teneur en polluants…), ce dans le but de caractériser les différents aquifères présents sur le tronçon. - le suivi de ces paramètres est ensuite réalisé par mesures et analyses chimiques tout au long du chantier. 3.3 Dissolution du gypse Le gypse est présent dans certaines couches sédimentaires présentes sur le tracé du Grand Paris Express : Masses et Marnes du gypse, Calcaire de Saint Ouen Sables de Beauchamp ou Marnes et Caillasses, essentiellement par exemple. La dissolution du gypse se produit lorsqu’il est soumis à un apport d’eau « non chargée en sulfate », ce phénomène peut entraîner soit une dégradation diffuse des caractéristiques mécaniques d’un horizon géologique, soit la création de vides de dissolution accompagnés de décompressions des terrains susjacents et/ou dans le cas extrême, d’apparition de fontis. Ce phénomène fait l’objet de plusieurs Plans de Prévention des Risques sur l’ensemble de la région Ile-de-France. 3.3.1

Prise en compte des enjeux dès la conception

Enjeux d’un projet de type métro souterrain liés à la dissolution du gypse

Pour vérifier que les mesures mises en œuvre en cas d’injection notamment sont efficaces, des sondages de contrôle des traitements seront réalisés. Par ailleurs, une méthode observationnelle sera mise en place dans ces zones, comme décrit au chapitre 2.1.3. 3.4 Retrait et gonflement des argiles L’argile voit sa consistance se modifier en fonction de sa teneur en eau. Ces variations de consistance s’accompagnent de variations de volume, dont l’amplitude peut s’avérer très importante. Les variations de volume générées par le retrait des argiles provoquent des tassements qui se manifestent par des désordres sur les ouvrages. A contrario, le phénomène de gonflement peut provoquer des soulèvements ou des sur-contraintes (pression de gonflement sous un radier de gare par exemple). En général, ces phénomènes se produisent à proximité de la surface, où la teneur en eau des argiles est soumise à de fortes variations, liées à la météorologie (périodes de sécheresse notamment), mais aussi à la végétation (système racinaire) ou à l’activité humaine (imperméabilisation des surfaces, pompages ou arrosages…). Ce phénomène fait l’objet de Plans de Prévention des Risques en Île-de-France 3.4.1

Les enjeux sont de deux natures : • La présence de zones décomprimées ou de vides dans le sous-sol préexistants sont potentiellement à l’origine des mêmes phénomènes que les carrières souterraines d’origine anthropique (voir détail au chapitre 3.1.2). De plus dans ce cas particulier, la réalisation d’un projet de type métro souterrain est susceptible d’activer ou de réactiver le phénomène de dissolution du gypse de par la modification éventuelle du régime d’écoulement des nappes d’eau souterraines dans les zones marquées par la présence de gypse. La création de ces vides pouvant avoir des impacts sur les travaux en cours de réalisation (arrêt du tunnelier, adaptation des méthodes constructives au niveau des gares) comme sur le bâti situé dans la zone d’influence hydrogéologique du projet. • La difficulté des zones de dissolution de gypse résidant le caractère aléatoire de sa répartition, et dans la difficulté de localiser avec certitude leur étendue.

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Les impacts d’un projet de type métro souterrain sur le retrait/gonflement des argiles

Un projet de type métro souterrain est susceptible d’activer ou de réactiver le phénomène de retrait/gonflement des argiles en modifiant le régime d’écoulement des nappes d’eau souterraines. Par ailleurs, la réalisation de terrassements à ciel ouvert est susceptible d’exposer des argiles aux aléas météorologiques alors qu’elles étaient jusqu’à présent protégées, favorisant également leur retrait/gonflement. Le projet du Grand Paris Express recoupe plusieurs formations géologiques argileuses considérées comme fortement sensibles. On citera en particulier les Argiles vertes et les argiles plastiques. 3.4.2

Prise en compte des enjeux dès la conception

Le projet du Grand Paris Express traverse les argiles vertes à l’est de Paris; les gares traversent cette couche et s’ancrent plus

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sondages carottés, 690 sondages pressiométriques, 180 sondages destructifs, menant ainsi l’espacement moyen entre sondages à environ 150 m, et 5 à 8 sondages par gare.

bas. Par conséquent, du fait des méthodes constructives retenues (parois moulées), les argiles vertes ne sont jamais exposées aux intempéries météorologiques. L’impact lié au retrait/gonflement des Argiles Vertes est donc quasi nul. Quant à la partie courante, le tunnel s’inscrit systématiquement sous cette couche, l’impact est donc également nul. Concernant la couche d’Argiles Plastiques qui se situe en profondeur, pour la partie en section courante située entre Arcueil-Cachan et Fort d’Issy, le tunnel s’inscrit partiellement dans cette couche. Comme mentionné précédemment (cf. 3.2.1), la réalisation du tunnel au tunnelier perturbe peu les conditions hydriques des Argiles Plastiques : le phénomène de retrait/gonflement des Argiles Plastiques est donc quasi nul pour cette partie. Concernant les gares ancrées dans les Argiles Plastiques, lors des terrassements en phase chantier, des précautions particulières pourront être mises en place en cas d’intempéries météorologiques (systèmes de drainage et de collecte, protection par des masques ou des écrans d’étanchéité ou des membranes, par exemple), le but étant de limiter l’intrusion d’eau dans la fouille et ainsi minimiser le gonflement des Argiles. Pour la phase définitive, les radiers des gares seront dimensionnés pour reprendre les efforts de gonflement des Argiles Plastiques. 3.4.3

Le projet Grand Paris Express ayant un développé important en termes de tracé et un planning d’études contraint, six entreprises de travaux de forages mènent à bien ces investigations, avec en moyenne 20 à 30 machines par mois le long des 150 km de tracé. Ces investigations font l‘objet d’un contrôle réalisé par l’Assistant à Maîtrise d’Ouvrage en Géotechnique de la SGP afin de garantir la qualité et l’homogénéité de ces dernières. Les reconnaissances engagées par la Société du Grand Paris sont en quantité importante au regard de la phase d’études à laquelle se situe le projet (pour rappel l’EUROCODE 7 EN 1997-2 de septembre 2007 « Calcul géotechnique » - annexe B relative aux ouvrages linéaires - préconise des sondages espacés de 20 à 200 m pour la phase finale de conception, soit la phase projet). Le but de ces reconnaissances conséquentes menées dès la phase G11 phase 2, est de permettre de stabiliser le modèle géologique, hydrogéologique et géotechnique le long du tracé au plus tôt afin de statuer sur les méthodes constructives. Il est reconnu que l’occurrence de désordres et accidents graves en travaux souterrains est inversement proportionnelle à la quantité et à la qualité de reconnaissances engagées lors des phases d’études. La Société du Grand Paris a donc, vu l’échelle du réseau, décider d’engager d’importantes investigations géotechniques dès les phases amont, ceci dans le but de maîtrises les risques.

Méthodes de suivi des effets des mesures

Au vu des mesures de réduction mises en œuvre dans la conception du projet, ainsi que de l’ampleur en conséquence très limitée du phénomène concerné, il n’y a pas de disposition spécifique à mettre en place dans le cadre du suivi des mesures. L’infrastructure du métro, comme toute infrastructure, fera l’objet au cours de son exploitation d’un suivi régulier, permettant de la maintenir efficacement.

La maîtrise des risques est au cœur de l’organisation des études via de multiples regards partagés sur les sujets techniques majeurs (bâti, géotechnique, méthodes constructives). La Société du Grand Paris a l’objectif d’étendre cette culture de la maîtrise des risques à l’ensemble de ses partenaires à venir (Conduite d’opération, Maîtres d’Œuvre, Entreprises).

4. LES CAMPAGNES DE RECONNAISSANCES En fonction des premiers éléments issus de la phase documentation et du profil en long préliminaire de projet, une campagne de reconnaissances géotechniques de type G11 au sens de la norme NF P 94-500 sur les missions d’investigations géotechniques (cf. titre 2.1.2) a été définie et réalisée en 2012 et 2013. Ce sont les problématiques rencontrées le long du tracé qui ont dicté le type de reconnaissances à effectuer. Au total, pour un tracé de 150 km et comprenant 57 gares (partie sous Maîtrise d’Ouvrage SGP), la campagne d’investigations géotechnique a compris : 385 sondages carottés dans lesquels ont été prélevés environ 1 500 échantillons intacts pour la réalisation d’essais en laboratoire; 278 sondages pressiométriques avec un essai pressiométrique tous les 1,5 m ; 19 forages destructifs avec enregistrement des paramètres de forage. La profondeur des sondages varie de 20 à 92 m de profondeur, avec une profondeur moyenne s’établissant aux alentours de 45m. L’espacement moyen entre sondages s’établit à environ 350m. Au droit de chacune des gares, lorsque le contexte urbain le permettait, il a été réalisé 3 sondages carottés et 2 sondages pressiométriques. La quasi-totalité des sondages carottés et des forages destructifs ainsi que et plusieurs sondages pressiométriques ont été équipés en piézomètres, conduisant ainsi à un total d’environ 450 piézomètres répartis le long du tracé pour reconnaître et suivre les différentes nappes concernées par le projet. Un relevé mensuel de l’ensemble de ces piézomètres est prévu pendant toute la durée des études. Les campagnes de reconnaissances pour la mission G12 vont débuter mi-2013 et porteront les investigations à environ 760

Pour rappel, le planning des travaux de réalisation du Grand Paris Express a une amplitude de 14 ans (2016 et 2030) comprenant 200 km de linéaire (principalement en tunnel) et 70 gares, avec des mises en service de tronçons s’échelonnant de 2020 à 2030 soit en moyenne 5 à 7 gares par an. Le Grand Paris Express est un projet d’envergure de par de nombreux aspects ; les investigations géotechniques entreprises en sont un ; elles sont à la hauteur de la volonté de la Société du Grand Paris de maîtriser les risques techniques.

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Innovations françaises françaises en en géotechnique: géotechnique: les les projets projets nationaux nationaux de de recherche recherche Innovations French Innovations in Geotechnics: the National Research Projects French Innovations in Geotechnics: the National Research Projects F. Schlosser; C. Plumelle; R. Frank, A. Puech, H. Gonin, F.Rocher-Lacoste, B. Simon Schlosser F., Plumelle C., Frank R., Puech A., Gonin H., Rocher-Lacoste F., Simon B. Comité Français de Mécanique des Sols et de Géotechnique Comité Français de Mécanique des Sols et de Géotechnique

C. Bernardini Bernardini C. Institut de Recherche Expérimental en Génie Civil (IREX) Institut de Recherche Expérimental en Génie Civil (IREX)

RÉSUMÉ : Un grand intérêt a été porté en France aux expérimentations en vraie grandeur en génie civil dès le début des années 1960 pour étudier le comportement des ouvrages et le mécanisme de nouveaux procédés. Cela a conduit un ingénieur des Ponts et Chaussées, M. Martin, à imaginer le concept novateur du Projet national (PN) de recherche expérimentale vers la fin des années 1970. L’originalité réside dans le fait que la majorité du financement est fournie par les participants eux-mêmes, sous forme de cotisations et surtout d’apports en nature (temps passé, essais, mise à disposition de matériel, de sites expérimentaux, etc.), le ministère concerné ne fournissant que 15 à 20 % du budget total. Le premier PN, Clouterre (1980-1985) sur le clouage des sols en soutènement, a été suivi de 30 PN en génie civil dont 7 en géotechnique. L’IREX (Institut de Recherche EXpérimentale en génie civil), organisme de gestion des PN, a été créé en 1989. On présente ici les débuts et la procédure des PN, illustrés par 5 PN en géotechnique. ABSTRACT: Full scale experiments have been considered of a great interest in French civil engineering since the 60’s beginning for studying structures behavior and new techniques mechanism. At the end of the 70’s the innovative concept of French experimental research project (FRP) was founded by the French civil engineer, M. Martin. The originality is that 80 to 85% of the funding is provided by the project members in the shape of subscriptions and contributions in kind (research time, experimental site, a.s.o.), the rest being financed by the ministry. The first project has been Clouterre (1980-85) on soil nailed retaining walls and since that time 30 projects in civil engineering have been carried out, which 7 in geotechnical engineering. The management organization IREX for these projects has been created in 1989. The paper presents the FRP organization illustrated by 5 projects in geotechnical engineering. MOTS CLÉS: recherche, projet, innovation, instrumentation, modèle physique et numérique, expérimentation en vraie grandeur. KEYWORDS : research, project, innovation, instrumentation, physical and numerical model, full scale experiment.

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INTRODUCTION.

Le comportement des sols est complexe et aucune théorie ne peut permettre de calculer correctement les contraintes et les déformations d’un sol sous une sollicitation quelconque. Ainsi le squelette d’un sol n’est ni élastique, ni même élasto-plastique. En outre le couplage eau-squelette est en général difficile à appréhender. Enfin, en dépit du remarquable développement de l’informatique, il n’a pas été possible d’obtenir un ensemble de relations entre contraintes et déformations représentant correctement le comportement d’ensemble d’un sol et utilisable en pratique. Toutes les théories ne sont qu’approchées. L’approche expérimentale du comportement des sols reste ainsi un élément primordial, notamment pour vérifier la validité d’une théorie. Les lois de la similitude de Mandel (1961) avaient déjà montré la limitation des modèles réduits en sable, sous sollicitation statique, par suite de l’effet d’échelle, ce qui a progressivement conduit au développement des centrifugeuses en géotechnique. Par ailleurs le développement important, depuis des dizaines d’années, des moyens de mesure a permis d’étudier non seulement certains aspects du comportement des ouvrages géotechniques en service, mais également de développer des ouvrages expérimentaux en vraie grandeur qui ont contribué à une grande amélioration des connaissances. En France, c’est J. Kerisel (1962) qui a réalisé un premier type d’ouvrage expérimental en vraie grandeur sur le comportement des pieux. Après avoir effectué sur le pont de Maracaibo au Venezuela, le premier essai de chargement de pieu en mesurant séparément l’effort en pointe et l’effort total en tête, il construit sur le site sableux de Saintt-Rémy-lès-Chevreuse une station d’essai de grandes dimensions où des pieux sont foncés

dans une grande et profonde cuve en béton remplie de sable compacté. Il y mesure séparément l’effort en pointe au cours de l’enfoncement et montre qu’il varie au début linéairement jusqu’à une profondeur d’environ trois fois le diamètre du pieu, puis reste constant au-delà. Ce résultat, maintenant bien connu, a largement contribué au changement du calcul de la résistance de pointe des pieux par rapport aux théories antérieurement appliquées. En France, un autre ouvrage expérimental en vraie grandeur fut réalisé toujours sur le site de Saint-Rémy-lès-Chevreuse par Tcheng (1975) sur la station du CEBTP afin d’étudier des grands massifs de sable mis progressivement en état de poussée ou de butée. L’élément principal de la station était un écran métallique très rigide, de 5 m de large et 3 m de hauteur, comportant dans sa partie centrale six cellules de mesure encastrées permettant d’y mesurer les composantes verticales et horizontales des contraintes. Il était suspendu au moyen de huit vérins hydrauliques et, à l’aide d’un système d’asservissement, il pouvait tourner autour d’un axe proche de la base et se translater horizontalement. Deux sables furent utilisés : le sable homométrique de Fontainebleau et le sable de Loire de granulométrie étalée. Les résultats furent intéressants, notamment sur les écarts entre la théorie et la réalité, mais ils montrèrent également les difficultés liées à une telle expérimentation, notamment l’état initial (K0) qui dépend du compactage et varie beaucoup du haut vers le bas de l’écran. À partir du milieu des années 1960, le Laboratoire central des ponts et chaussées a développé, en coopération avec les laboratoires régionaux des ponts et chaussées, des recherches sur les remblais sur sols compressibles (1973), la stabilité des pentes (1976), les fondations profondes et les nouveaux ouvrages de

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soutènement. Dans chacun des cas, un ou plusieurs ouvrages expérimentaux en vraie grandeur étaient construits spécialement pour cette recherche. Pour la stabilité des pentes, un versant naturel instable avait été dédié à la recherche et largement instrumenté, puis suivi pendant plusieurs années. Des recherches sur la nouvelle technique de soutènement française de la Terre Armée, inventée par Henri Vidal en 1963, furent entreprises pour aboutir aux Recommandations et Règles de l’art (1979) rédigées conjointement par le LCPC et le Service d’Études Techniques des Routes et Autoroutes (SETRA). Un mur expérimental en Terre Armée fut construit en 1968 par le Service des Ponts et Chaussées du département de l’Eure et instrumenté par le LRPC de l’Ouest parisien. Il permit de montrer pour la première fois que l’effort de traction dans les armatures n’était pas maximal au parement, mais à une certaine distance à l’intérieur du mur (Figure 1).

une première expérience: les tunnels et la technique du clouage des sols pour les soutènements. En dépit d’un certain scepticisme au début, ces deux projets géotechniques, réalisés entre 1985 et 1989, furent un succès. Ainsi le PN Clouterre sur le clouage, qui débuta en 1986 pour 4 années de recherches, a comporté 21 membres (7 organismes publics, 3 maîtres d’ouvrage publics et privés, 11 entreprises). Son budget fut de 3,15 M€ dont 15% apportés par la DAEI et 85% financés directement par les 21 membres avec les cotisations et les prestations en nature. La gestion du projet fut assurée par un des partenaires : le CEPTP, qui mit à la disposition du projet son site expérimental de St Rémy lès Chevreuse. Après ces deux premiers projets nationaux, il fut reconnu nécessaire d’avoir une structure vraiment adaptée au caractère collectif des PN pour en assurer la gestion, le suivi et également la diffusion des résultats. C’est ainsi qu’a été créée, en 1989 et de façon conjointe par le ministère de la Recherche et le ministère de l’Équipement, l’Institut pour la Recherche et l’Expérimentation en génie civil (IREX). 3 LA PROCÉDURE DES PROJETS NATIONAUX DE RECHERCHE.

Figure 1. Expérimentation en vraie grandeur du mur en Terre Armée d’Incarville (1968).Evolution de la force de traction dans les armatures instrumentées d’un lit situé à 3m de profondeur.

Toutes ces recherches du LCPC et des Laboratoires Régionaux étaient financées par le ministère de l’Équipement dont dépendait le LCPC. Il n’y avait alors en France aucune centralisation de la recherche en génie civil. Les universités n’étaient pas associées à ces recherches et les grandes entreprises, comme les grands services de l’état (SNCF, EDF, etc.), effectuaient dans ce domaine leurs propres recherches. C’était l’époque du début des autoroutes financées par l’État, étudiées et construites par les Services des Ponts et Chaussées. 2 LA NAISSANCE DES PROJETS NATIONAUX DE RECHERCHE EN GÉNIE CIVIL. C’est à un ingénieur des Ponts et Chaussées, Michel Martin, alors en service à la Direction des Affaires Étrangères et Internationales (DAEI) du ministère de l’Équipement, que revient l’idée des Projets Nationaux sur des recherches expérimentales en génie civil, développée au tout début des années 1980. Il s’agissait d’une part de permettre des projets de recherche d’une assez grande ampleur, d’autre part et surtout de rassembler sur un thème de recherche le plus grand nombre possible de participants à la fois publics et privés. Le principe consistait à demander aux participants une cotisation financière pour chaque année de recherche, puis à leur permettre de participer au financement des recherches sous la forme d’apports en nature (temps passé, essais, mise à disposition de matériel, etc.) et enfin à fournir une subvention financière du ministère de l’Équipement égale à 15% ou 20 % du montant total du projet. Deux thèmes furent choisis pour réaliser

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La procédure actuelle, qui vise à développer la recherche appliquée et expérimentale en génie civil, a été initiée vers 1990 conjointement par les ministères de la Recherche et de l’Equipement sur proposition d’un Conseil d’Orientation de la Recherche en Génie Civil (CORGEC) comprenant des représentants du monde de la recherche et du génie civil Elle comprend tout d’abord la validation, par la Direction de la Recherche du ministère de l’Équipement, d’un thème de recherche proposé par la profession sur la base d’une étude de faisabilité réalisée par un groupe d’experts animé par l’IREX. Cette étude est rémunérée à l’aide d’une subvention du ministère de l’Équipement, après avis d’un Comité d’orientation du génie civil et urbain regroupant des chercheurs de l’Université et des Centres techniques de l’État, ainsi que des représentants de la profession. À la suite de cette étude, l’IREX monte un dossier détaillé du Projet National comprenant : le programme de recherche avec ses expérimentations, la liste de ses partenaires publics et privés, le planning qui s’étale en général sur quatre ans, le coût du projet et son financement (cotisations, apports en nature, subvention du ministère de l’Équipement entre 15 et 20%). Il est à noter que le dossier doit toujours comprendre au moins un maître d’ouvrage qui accepte de prendre totalement ou partiellement à sa charge une expérimentation en vraie grandeur ou une instrumentation très complète d’un ouvrage. Il est également demandé de prévoir un poste de valorisation du projet pour réaliser une synthèse des résultats, puis de la publier sous forme de recommandations ou de guide. La plupart du temps, une version en anglais est publiée. Les avancées techniques les plus marquantes font par ailleurs l’objet de présentations dans les congrès internationaux. Les Projets Nationaux ont couvert une large gamme du génie civil: 1) les matériaux, essentiellement les divers types de béton ; 2) la géotechnique avec principalement les fondations ; 3) les procédés de construction ; 4) la réhabilitation et la maintenance ; 5) le développement durable En 2009, à l’occasion de l’anniversaire des 20 ans de l’IREX, un document de synthèse sur les Projets Nationaux a été publié, intitulé « 20 ans de recherches appliquées et d’expérimentations en génie civil ». Il donne, en 4 à 6 pages pour chacun des 26 Projets Nationaux, une description du projet et de ses retombées. Nous nous intéresserons ici aux PN suivants qui se classent dans la géotechnique ou qui s’y rattachent, soit :

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 CLOUTERRE : technique du clouage des sols en soutènement.  FOREVER :technique des groupes et réseaux de micropieux  VIBROFONÇAGE : vibrage des pieux et des palplanches.  ASIRI : amélioration des fondations par inclusions rigides  SOLCYP : comportement des pieux sous charges cycliques On donne ci-après deux tableaux qui montrent l’un la répartition des partenaires, l’autre les montants financiers respectifs de ces 6 PN.

3. Depuis l’origine, la Fédération nationale des travaux publics est acteur des projets nationaux et les entreprises, bien que concurrentes, ont su unir leurs efforts et trouver des dénominateurs communs pour faire évoluer les doctrines techniques, les référentiels et les règlements, mais aussi pour utiliser les progrès ainsi obtenus au développement de leur activité à l’international. 4. Les projets nationaux ont permis aux ingénieurs publics et privés de travailler ensemble dans des domaines de recherche et de s’apprécier, alors qu’auparavant de telles opportunités n’étaient que très occasionnelles.

Tableau 1. Répartition des partenaires dans des projets nationaux géotechniques.

4 4.1

LE PROJET NATIONAL CLOUTERRE Objectif et caractéristiques du projet.

Le but de ce Projet national était de promouvoir le clouage des sols, notamment pour les ouvrages de soutènement permanents, grâce à une connaissance approfondie du procédé, à la détermination des limites du procédé, à la mise au point de méthodes de dimensionnement fiables et à la rédaction de recommandations. Tous ces points étaient à développer en s’appuyant sur des expérimentations en vraie grandeur. En fait, quelques années après la publication des Recommandations CLOUTERRE 1991, il s’est avéré nécessaire de compléter les résultats du PN CLOUTERRE I en effectuant des recherches sur les murs et autres ouvrages en sol cloué, en particulier de développer une méthode de dimensionnement aux états limites de service (ELS) à partir de calculs aux éléments finis. Ce sera le Projet national CLOUTERRE II dont les recherches ont été effectuées de 1995 à 1999.

Tableau 2. Montants financiers de projets nationaux géotechniques.

Quelques aspects particuliers de ces Projets Nationaux sont par ailleurs à noter : 1. Compte tenu de la création récente de l’Agence Nationale de la Recherche (ANR), pilotée par le ministère de la Recherche, plusieurs projets ont bénéficié d’une subvention de cet organisme pour des recherches à effectuer en laboratoire, alors que les recherches plus orientées vers les expérimentations ont fait l’objet d’une subvention du ministère de l’Équipement. (voir tableau 2). Toutefois la collaboration entre les différents partenaires n’en a pas été modifiée et est toujours restée très féconde. 2. Bien que le mot « national » pourrait laisser penser qu’il soit fait exclusivement appel à des partenaires français, plusieurs projets nationaux en géotechnique ont eu des partenaires étrangers. Ainsi le ministère des Transports du Québec a été partenaire dans le projet CLOUTERRE, la Federal Highway Administration (États-Unis) et l’université de Canterbury (Nouvelle-Zélande) ont été partenaires dans le projet FOREVER. En outre, dès 1991 ce processus des projets nationaux a intéressé d’autres pays : le ministère fédéral de la Recherche au Canada en 1991, puis son homologue en Chine (1992) et, plus récemment, une mission Japonaise en France.

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Figure 2. Les trois phases de construction d’un mur de soutènement par clouage d’un sol en place.

Comme le montre la figure 2, la technique de soutènement par clouage se situe dans le prolongement de la Terre Armée dont les premiers grands ouvrages furent réalisés en 1968-1969 pour l’autoroute entre Nice et Menton, dont le mur du Peyronnet de 23 m de hauteur qui n’a pas bougé depuis. Cependant la construction, à l’inverse de la Terre Armée, se fait du haut vers le bas, ce qui change bien des choses et complique la réalisation. En particulier la phase de terrassement, à la base de la partie déjà construite du mur, peut, si elle est de hauteur trop importante et/ou laissée en place lors d’un arrêt de chantier de plusieurs jours, conduire à une rupture. Comme indiqué précédemment Le PN Clouterre I avait 22 partenaires dont le ministère des Transports du Québec. Il s’est déroulé de 1986 à 1990 et son budget global a été de 3.150 000 € dont 15 % financés par le ministère de

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rupture locale se propage jusqu’à la surface conduisant à une rupture globale et interne du mur.

l’Équipement, le solde étant apporté par les partenaires sous la forme de cotisations et d’apports en nature. 4.2

Les expérimentations en vraie grandeur de CLOUTERRE I.

Sur le site expérimental du CEBTP à Saintt-Rémy-lèsChevreuse, il a été possible de réaliser trois expérimentations en vraie grandeur de murs en sol cloué, construits dans des massifs de sable de Fontainebleau rapporté aux caractéristiques bien contrôlées. Ils sont sommairement décrits ci-après. 4.2.1 Mur n° 1 en sol cloué poussé jusqu’à la rupture. Ce mur de 7 m de hauteur en sable a été réalisé par phase d’excavation de 1 m de hauteur avec des clous scellés de 6 à 8 m de longueur et dotés d’une certaine résistance à la flexion car constitués de tubes. L’ouvrage avait été calculé avec un coefficient de sécurité global suffisamment faible (F = 1,1) pour pouvoir être rompu facilement en saturant progressivement le sol à partir de la tête du mur, ce qui diminuait la cohésion apparente du sable et augmentait son poids volumique total. Grâce à l’instrumentation très complète mise en place, il a été possible d’effectuer de nombreuses mesures (tractions dans les clous, déplacements du parement, déformation du massif en sol cloué, etc.). De plus, la rupture n’ayant pas été totale, le parement s’étant enfoncé et bloqué dans le sol de fondation, l’excavation du mur a permis une investigation très complète du comportement de l’ouvrage à la rupture (Figure 3).

Figure 3. Observations lors de l’excavation du mur en sol cloué après sa rupture (1ère expérimentation en vraie grandeur au CEBTP)

En particulier la flexion des clous au voisinage de la rupture entraîne l’existence d’une zone de cisaillement dans le sol autour de la ligne des points de traction maximale dans les clous, ainsi qu’en règle générale un aspect non brutal mais ductile de la rupture du mur. 4.2.2. Mur en sol cloué avec étude de la phase d’excavation. L’objectif de ce mur expérimental n°2 du CEBTP fut d’étudier la stabilité, aussi bien locale que globale, d’un massif en sol cloué en phase d’excavation. Pour ce faire, un mur en sol cloué de 3 m de hauteur à été construit puis poussé à la rupture par augmentation de la hauteur d’excavation en pied de mur de 1 m à 3 m. À la première passe (1 m de hauteur d’excavation), l’excavation, comme le mur, était stable. À la deuxième passe (2 m de hauteur d’excavation), une rupture localisée s’est produite suivie d’une stabilisation par formation d’une voûte, mais le mur est globalement resté stable. À la troisième passe (3m de hauteur d’excavation), l’effet de voûte se détruit et la

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Figure 4. Stabilité et rupture de la phase d’excavation dans le mur en sol cloué n°2 du CEBTP.

4.2.3. Mur n° 3 en sol cloué avec rupture par insuffisance de longueur des clous. La troisième expérimentation d’un mur en sol cloué au CEBTP, de 6m de hauteur, a permis d’étudier le mode de rupture par insuffisance de longueur des renforcements. Il a été mis en place des clous télescopiques dont on pouvait réduire la longueur. La rupture s’est produite lorsque a été atteinte une répartition de clous très courts à la base du mur et augmentant progressivement de longueur vers le haut du mur .Cette disposition a imposé la forme de la surface de glissement correspondant à une rupture intermédiaire entre le mode par défaut d’adhérence et le mode par rupture externe. 4.3

Principaux résultats de Clouterre I.

Le mur n°1 a montré la forme de la ligne des tractions maximales dans les clous, laquelle n’évolue pas jusqu’à l’initiation de la rupture qui est progressive, ainsi qu’une certaine mise en flexion des clous au voisinage de la rupture. Le mur n°2 a montré que la stabilité du mur durant sa construction était liée au développement d’un effet de voûte lors des phases d’excavation, ce qui a notamment donné des informations sur la limite du procédé. Le frottement sol/clou a quant à lui fait l’objet d’études approfondies tant expérimentales que théoriques, avec comme dans la Terre Armée la notion de coefficient de frottement apparent * liée à une dilatance en partie empêchée de la partie granulaire du squelette du sol. Une part importante des recherches a été consacrée à la mise au point d’une méthode dimensionnement à l’état limite ultime (ELU). Le choix a été porté sur une méthode à la rupture utilisant des surfaces de rupture circulaires, notamment calée sur le mur en vraie grandeur n°1. Il a notamment été développé une méthode dite du multicritère (Schlosser, 1982) qui permet de déterminer le torseur (Tn, Tc, M) des efforts au point de traction maximale dans un clou. Elle fait intervenir des critères de rupture portant sur les constituants et les interactions entre constituants : - interaction de frottement latéral sol/clou :   qs - interaction de pression latérale sol/clou : p  pmax - matériau constitutif du clou :   k (cission) Cela conduit à quatre critères compte tenu de l’assimilation des clous à de poutres. Il en résulte dans le plan (Tn, Tc) des efforts de traction et de cisaillement un domaine de stabilité qui permet de déterminer l’effort résultant maximal (Figure 5) Le multicritère permet de prendre en compte un effort de cisaillement dans les clous, qui est souvent négligé dans le dimensionnement des murs en sol cloué, mais qui devient prépondérant dans le clouage vertical utilisé pour la stabilisation des pentes. Cette méthode de dimensionnement fut la première en mécanique des sols à utiliser le calcul semi probabiliste avec coefficients de sécurité partiels et coefficients de pondération sur les actions, ce qui est maintenant devenu la règle dans les Eurocodes.

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En outre un chapitre a été consacré au comportement et à la justification du parement à partir d’instrumentations d’ouvrages en service, notamment les murs de l’autoroute A 12 au sud-ouest de Paris. 4.5.2.2. Méthodes de calcul aux déplacements. La banque de données de Clouterre 1991 sur les mesures des déplacements des murs réels en sol cloué a été complétée et une méthode semi-empirique a été mise au point à partir de ces résultats. Des méthodes générales à partir des éléments finis (logiciels CESAR, Plaxis) et des différences finies (logiciel FLAC-2D) ont été développées. Elles font appel à des modélisations en bidimensionnel : celle où les clous sont modélisés par des « plaques équivalentes avec interfaces planes » pour les éléments finis, celle des « clous équivalents avec fonctions de transfert de charge » pour les codes aux différences finies. Leur validation a été faite en comparant les résultats des calculs avec des mesures réalisées sur des ouvrages en vraie grandeur, construits dans des sols aux propriété connues, instrumentés et suivis dans le temps, depuis la construction jusqu’à la mise en service et éventuellement la rupture. Les murs expérimentaux de Clouterre I ont à ce sujet constitué une base exceptionnelle d’ouvrages de référence. En plus des paramètres classiques d’élasticité et de résistance des sols, il faut ajouter l’ange de dilatance  ainsi que les paramètres relatifs aux clous, au parement et à leurs interactions avec le sol. La figure 6 donne les évolutions du déplacement horizontal h en tête de parement du mur Clouterre n°1 et la comparaison avec les valeurs mesurées. Globalement et par comparaison avec les calculs antérieurs effectués (Shaffiee, 1986), on constate une bonne prédiction des valeurs mesurées, mais il est recommandé de réaliser une étude de sensibilité aux paramètres mécaniques pour s’assurer de la validité des résultats

Figure 5 Domaine de stabilité dans le plan (Tn, Tc) et détermination de l’effort maximal T.

La déformation des murs en sol cloué, avec notamment les déplacements en tête a également fait l’objet de nombreuses instrumentations tant sur les murs expérimentaux que sur des ouvrages en service. Le déplacement horizontal en tête d’un mur en sol cloué vertical de hauteur H est ainsi compris entre H/1 000 et 3H/1 000 suivant la valeur du coefficient de sécurité. 4.4

Publications de Clouterre 1

Le PN a fait l’objet de 50 rapports internes et de publications à la fois en France et à l’étranger. La publication la plus importante fut les Recommandations Clouterre 1991 pour la conception, l’exécution et le contrôle des soutènements réalisés par clouage des sols. Ce livre comprend sept chapitres et, après avoir été traduit en anglais, a été édité à 10 000 exemplaires par la FHWA (Federal Highway Administration) aux États-Unis, puis publié en commun par la FHWA et les Presses des ponts et chaussées dans le monde entier. Ces recommandations ont contribué à un large essor de la technique des soutènements en sol cloué et ont abouti à la norme PR-94270 qui constitue la norme d’application française de l’Eurocode 7 pour ce qui concerne à la fois les ouvrages de soutènement en sol cloué et en sol renforcé. 4.5

Le Projet national Clouterre II.

4.5.1. Organisation du P.N. Développé à la suite de Clouterre I de 1993 à 1997, le Projet National CLOUTERRE II a eu 19 partenaires dont la FHWA, ce qui a constitué le premier exemple d’un partenaire étranger participant à un PN. Le coût total de ce PN s’est élevé à 1 579 190 € dont une subvention de la DRAST de 281 708 € représentant 17,8% du budget total, le solde étant fourni par les partenaires (cotisations et apports en nature).

Figure 6. Evolutions calculées du déplacement horizontal en tête du parement et comparaison avec les mesures (mur Clouterre n°1).

4.5.2.3 Autres recherches.  Étude de l’effet du gel-dégel. Une instrumentation effectuée sur un mur en sol cloué construit en montagne en 1982 et ayant subi des déplacements importants du parement lors d’une période de gel a permis d’étudier les mécanismes du gel-dégel et de mettre au point une prise en compte des effets du gel dans un mur en sol cloué.  Dimensionnement sous séisme. Les murs en sol cloué sont, comme les murs en Terre Armée (Kobayashi et al., 1996), des ouvrages souples qui résistent bien aux séismes. En règle générale, leur stabilité au séisme est analysée par un calcul à la rupture en utilisant la méthode pseudo-statique. Le cas des ouvrages mixtes où le mur en sol cloué est conforté en tête par des tirants précontraints nécessite de prendre en compte dans les calculs de stabilité des coefficients sismiques aux valeurs majorées.  Clouage (boulonnage) du front de taille des tunnels en terrain meuble. Utilisé depuis 1985, le boulonnage par barres

4.5.2. Les recherches du P.N. Clouterre II. Clouterre II a marqué une étape complémentaire dans la connaissance et le dimensionnement des ouvrages en sol cloué, l’accent ayant été mis sur les méthodes de calcul des déplacements, l’exécution, le comportement sous des sollicitations particulières (murs soumis au gel, aux séismes) et le comportement d’un ouvrage autre que les murs (front de taille renforcé par des clous dans les tunnels en terrain meuble). 4.5.2.1. Exécution des murs en sol cloué. Comportement et justification du parement. Il a été fait une mise à jour de la banque de données des essais de traction de clous de CLOUTERRE I, en particulier des abaques (qs, pl) donnant pour les différentes catégories de sols les valeurs de la contrainte de frottement limite qs de l’interaction sol/clou en fonction de la pression limite pl au presiomètre. L’exécution du parement a fait l’objet d’ajouts en particulier sur le drainage.

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scellées est utilisé pour stabiliser les parements ou le front de taille en pleine section d’un tunnel en construction. Dans ce second cas les barres sont en matériau composite de fibres de verre et de résine afin de pouvoir être facilement détruites à l’excavation. La stabilité du front renforcé est calculée par une analyse à l’équilibre limite ou en utilisant la théorie du calcul à la rupture. Les calculs en déformation pour prendre en compte ce clouage dans les déplacements du terrain et les soutènements sont de 3 types : 1. Modélisation de chaque barre et calcul en 3D. 2. Approche par homogénéisation de l’ensemble terrain et barre et calculs en 2D axisymétriques. 3. Simuler l’effet du clouage par une pression appliquée au front de taille et calculs en 2D axisymétriques. 4.5.3. Publications de Clouterre II. Les travaux du Projet National CLOUTERRE II ont fait l’objet de 22 rapports internes et d’un livre Additif 2002 aux recommandations CLOUTERRE 1991, édité par les Presses des Ponts. Il comprend 8 chapitres élaborés et mis au point par un comité de rédaction de 12 personnes. 4.6

Retombées des Projets Nationaux CLOUTERRE I et II

On peut affirmer sans pouvoir vraiment le quantifier que ces deux PN ont contribué en France à un grand essor des murs en sol cloué en tant qu’ouvrages permanents, permettant de ce fait une économie importante par rapport à des murs plus classiques. On peut citer par exemple les murs en sol cloué autour de certaines piles du viaduc de Millau. Conçus initialement comme des ouvrages provisoires, ces murs ont été, au moment de la remise en état des lieux à la fin de la construction du viaduc, transformés en ouvrages permanents et inclus dans l’ensemble du processus de suivi des divers éléments du viaduc, mais avec une démarche du type méthode observationnelle. L’économie par rapport à de nouveaux ouvrages de soutènement en béton armé a été substantielle. Par ailleurs il est intéressant de noter le classement en « ouvrage de référence » en 1998 par le comité IVOR (Innovations Validées sur Ouvrages de Référence) des murs de soutènement en sol cloué de l’autoroute A12 qui ont fait l’objet d’une instrumentation importante dans le cadre de CLOUTERRE II. A l’international, c’est incontestablement le Projet National CLOUTERRE I, avec la traduction anglaise des Recommandations CLOUTERRE 1991, qui a été à l’origine d’un fort rayonnement de la technique française, lequel a notamment conduit à la participation de l’Administration des Autoroutes Fédérales Américaines (FHWA) en tant que partenaire à CLOUTERRE II, puis plus tard au Projet national FOREVER. Il est intéressant de noter que le logiciel Talren, conçu et développé par Terrasol, a été et reste très largement utilisé dans de nombreux pays pour le dimensionnement des ouvrages en sol cloué (murs, talus et pentes). C’est ainsi .que les Recommandations CLOUTERRE 1991 .ont été traduites en coréen. Au tout début des années 90, la FHWA et le TRB (Transportation Research Board) des Etas Unis avaient organisé un « scanning tour » en Europe pour y connaître le développement du clouage. Ils furent très favorablement impressionnés par l’essor du clouage en France. De même que la Terre Armée a connu un développement remarquable aux Etats Unis, le clouage des sols y a eu un essor rapide et sans doute plus important à tel point que le bénéfice cumulé obtenu grâce à l’utilisation de cette technique a pu être estimé voici quelques années par l’administration américaine à plusieurs centaines de millions de dollars. A l’heure actuelle, le clouage des sols est utilisé dans la quasi totalité du monde, car il s’agit d’une technique simple, facile à mettre en œuvre et non protégée par des brevets.

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5 5.1

LE PN FOREVER SUR LES MICROPIEUX. Objectif et organisation

Un micropieu est un pieu de diamètre inférieur à 250 mm, le plus souvent foré et comportant une armature métallique centrale, souvent un tube, scellée dans un mortier ou un coulis de ciment. La capacité portante est essentiellement assurée par le frottement latéral micropieu/sol qui peut être d’autant plus important que le coulis a été injecté sous forte pression. On distingue 4 types de micropieux fonction principalement de la valeur de la pression d’injection du coulis : - Type I. Foré et tubé, équipé ou non d’une armature, rempli d’un mortier de ciment au tube plongeur. Tubage récupéré. - Type II. Foré, équipé d’une armature et rempli au tube plongeur d’un mortier ou coulis de ciment par gravité ou sous très faible pression. - Type III. Le plus souvent foré, équipé d’une armature et d’un système d’injection du coulis par tube à manchettes mis en place dans un coulis de gaine. L’injection est globale et unitaire avec une pression d’injection en tête supérieure ou égale à 1MPa. - Type IV. Identique au type III, mais l’injection y est répétitive et sélective à l’obturateur simple ou double Depuis de nombreuses années, les micropieux offrent un vase champ d’applications en groupe (ensemble de micropieux verticaux) ou en réseau (ensemble de micropieux inclinés). Ils sont d’abord utilisés pour la reprise en sous-œuvre des fondations, mais également pour les fondations d’ouvrages neufs en terrain difficile, pour la stabilisation des pentes et des talus ainsi que pour les soutènements, les tunnels et la protection de structures enterrées. Les réseaux de micropieux ont également de remarquables capacités de résistance aux actions sismiques. L’objectif du projet national FOREVER (FOndations Renforcées VERticalement) a été, grâce à un programme d’études et d’essais en vraie grandeur, de préciser le comportement de micropieux isolés, en groupe ou en réseau, puis d’établir des règles de l’art ainsi que des méthodes de dimensionnement permettant d’élargir leur champ d’application. Des groupes et réseaux expérimentaux ont été construits et instrumentés sur le site du CEBTP à St Rémy lès Chevreuse. La direction du PN comprenait un président, un directeur scientifique et un directeur technique. Le projet a eu 22partenaires et il s’est déroulé de 1993 à 2001. Son budget s’est élevé à 5 091 000 € dont 754 000 € de subvention de la DRAST et le solde en apport des partenaires (cotisations et apports en nature). Il est à noter que trois partenaires étrangers ont fait partie de Forever : la Federal Highway Administration (EtatsUnis), l’Université de Canterbury (Nouvelle Zélande) et la Polytechnic University de New York (États-Unis) 5.2

Groupes de micropieux. Résultats expérimentaux.

Il a été confirmé, à partir de nombreux essais réalisés par Forever en modèle réduit (chambre d’étalonnage, centrifugeuse) et d’un essai en vraie grandeur, que l’espacement S entre les micropieux d’un groupe dans du sable est l’un des paramètres les plus influents sur la capacité portante sous charge verticale. Le coefficient d’efficacité Ce , rapport entre la capacité portante moyenne d’un micropieu du groupe et celle du micropieu isolé, varie entre 0,59 et 2,2. Pour les mêmes essais, le nombre N de micropieux du groupe s’avère également être un paramètre influent : pour N10 Ce est compris entre 1,4 et 2,2. L’ordre d’installation des micropieux a également une influence. Ainsi, pour un groupe de 5 micropieux foncés dans un sable moyennement dense, la mise en place d’un 5ème micropieu au centre des 4 autres augmente sa capacité portante de 40%.

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En revanche la capacité portante sous charge horizontale d’un groupe de micropieux s’avère être semblable à celle d’un groupe de pieux. 5.3

Groupes de micropieux. Méthodes de calcul numériques

5.3.1. Programme GOUPEG En 1994 Maleki et Frank ont développé le programme GOUPEG pour les groupes de micropieux à partir du programme GOUPILLCPC de 1989 qui utilise les fonctions de transfert pour les chargement axiaux (courbes de mobilisation t-z pour le frottement latéral axial) et pour les chargements latéraux (courbes de réaction latérale p-y). Leur étude a consisté à introduire dans GOUPEG l’effet de groupe dans le cas de forces axiales. Il s’agit d’une méthode « hybride » dans laquelle on utilise les solutions en élasticité de Mindlin pour calculer de façon automatique les déplacements induits sur les pieux voisins et ainsi déterminer les facteurs type «y» (c'est-à-dire les déplacements z) pour corriger les courbes de mobilisation t-z du frottement latéral (et de la résistance en pointe q-zp). Le programme GOUPEG a été validé en comparant les coefficients d’interaction F obtenus avec les solutions bien connues en continuum élastique de Poulos et Davis (1990). 5.3.2. Interprétation des essais de Rueil Malmaison. Ces essais furent réalisés sur 4 micropieux verticaux : 1 micropieu isolé et 1 groupe de 3 micropieux espacés de 1m et tirés en traction. Ces micropieux étaient constitués de tubes d’acier de diamètre B = 89 mm avec une longueur libre de 14 m dans les alluvions et une longueur scellée de 5 m dans la craie sous-jacente (B = 125 mm). Ils étaient instrumentés en 8 sections avec un extensomètre amovible du LCPC pour déterminer le frottement le long du fût. Plusieurs calculs du chargement en traction ont été faits avec GOUPEG et chaque fois les lois de mobilisation du frottement latéral furent celles de Frank et Zhao. Pour l’interaction entre les micropieux et l’utilisation des solutions de Mindlin, un module d’Young E= 10 EM (EM module pressiométrique) a été pris.

Figure 7. Comparaisons entre les courbes de chargement en traction mesurées et calculées des micropieux du groupe. (Essai de RueilMalmaison)

La figure 7 donne les comparaisons entre les courbes de chargement mesurées en tête de chaque micropieu du groupe et les courbes calculées avec GOUPEG. suivant 2 hypothèses pour le frottement latéral limite (I. Valeur moyenne mesurée sur le micropieu isolé. II. Valeur moyenne mesurée sur le groupe). Les résultats sont satisfaisants sachant que la longueur libre du micropieu isolé est inférieure à 14m à cause d’une remontée de coulis. 5.3.3. Analyse de l’essai de chargement latéral à St Rémy. Le programme GOUPEG a été étendu à l’analyse des groupes de micropieux sous chargement latéral, toujours avec utilisation des équations de Mindlin. Il a ainsi permis d’étudier les chargements vertical et latéral d’essais en vraie grandeur sur le site expérimental en sable du CEBTP à St Rémy, comprenant des

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micropieux isolés et en deux groupes de 4 de caractéristiques : S/B = 2 ; D = 5m ; B = 10 cm ; ID = 0, 57. Les données nécessaires à l’analyse étaient les courbes « t-z » pour le frottement latéral, « q-z » pour la résistance en pointe et « p-y » pour la résistance latérale, courbes exclusivement tirés des essais pressiométriques effectués sur le site. Pour le calcul de l’effet de groupe (interaction pieu-sol-pieu), le module de cisaillement G du sol (intervenant dans les équations de Mindlin) a dû être évalué. La figure 8 montre la comparaison des résultats expérimentaux et des calculs GOUPEG pour le groupe de micropieux de type II (coulis mis par gravité). On constate que la méthode pressiométrique développée pour les pieux est également valable pour les micropieux et que, pour le groupe de micropieux, la tendance donnée par GOUPEG représente bien la réalité.

Figure 8. Comparaison des courbes effort-déplacement mesurées et calculées par GOUPEG dans le chargement horizontal d’un groupe de micropieux de type II.(Essai de St Rémy)

5.4

Réseaux de micropieux chargés verticalement.

Les recherches expérimentales sur les réseaux de micropieux, dans lesquels tous les micropieux sont inclinés et où des chargements verticaux ont été effectués, sont rares. Les nombreux essais réalisés dans FOREVER sur des réseaux chargés verticalement dans du sable ont permis d’analyser l’influence des paramètres suivants : espacement des micropieux, densité du sable, densité et enchevêtrement des micropieux. L’orientation des micropieux dans un réseau est caractérisée par deux angles : l’angle  de l’inclinaison du micropieu avec la verticale et l’angle , appelé angle d’enchevêtrement, entre le plan vertical contenant le micropieu et le plan vertical tangent au cercle horizontal centré au milieu de la fondation et passant par la tête du micropieu. Un réseau enchevêtré est caractérisé par des valeurs négatives de  (180°) qui permettent aux micropieux d’avoir des distances entre eux plus faibles qu’en tête conduisant à un plus grand confinement du sol entre les micropieux. Les premiers résultats expérimentaux ont été établis par Lizzi (1978) qui a comparé en modèle réduit au 1/10 les comportements d’un groupe et d’un réseau de 18 micropieux chacun. L’amélioration apporté par le réseau était donné par le coefficient d’efficacité Ce = 1,68 , rapport entre les capacités portantes du réseau et du groupe, ou Ce0 = 1,22 , rapport entre la capacité portante du réseau et la somme des capacités portantes des micropieux isolés verticaux. Le PN FOREVER a réalisé un ensemble de 20 essais sur des réseaux dans du sable (vraie grandeur, centrifugeuse, cuve, chambre d’étalonnage) en faisant varier les paramètres. Le premier résultat est la grande dispersion des valeurs du coefficient Ce0 (0,51 à 2,93) qui s’explique en partie par le mode de mise en place des micropieux : fonçage, forage, moulage. L’espacement relatif S/B n’apparaît pas comme un paramètre principal. La densité du sable n’a guère pu être étudiée car pour tous les essais l’indice de densité ID du sable était voisin de 0,5

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correspondant à un sable moyennement lâche. Quant à la densité de micropieux ou à leur nombre N, il apparaît qu’il faille un nombre minimum de micropieux par unité de volume pour conduire à un effet de groupe positif. En ce qui concerne l’orientation des micropieux et les valeurs des angles  et , il n’a là aussi pas été possible de dégager d’effet précis car trop peu d’essais permettaient de faire varier l’un de ces deux paramètres en gardant tous les autres constants. Il peut cependant être confirmé que l’inclinaison d’un micropieu isolé est défavorable pour sa capacité portante verticale comparée à celle du même micropieu en position verticale. Cependant il a été montré, sur les réseaux simples que constituent les chevalets, qu’un mécanisme spécifique aux micropieux inclinés se développe lors d’un chargement vertical, à savoir la mobilisation progressive d’une butée avec flexion sur les micropieux. Ce phénomène, qui a été également mis en évidence dans les études numériques effectuées au CERMES, peut conduire à une capacité améliorée par rapport à celle du groupe équivalent. Les essais en cuve effectués au Laboratoire 3S de Grenoble ont utilisé des réseaux ayant un nombre nettement plus élevé de micropieux (N = 18) et un meilleur enchevêtrement ( < 0° et  > 180° avec des intersections de micropieux) comme le montre la figure 9. Dans le cas d’un réseau quasi cylindrique (défini par  = 20° et  = - 30°/ 210°), il est observé un effet positif sur la capacité portante par rapport au groupe équivalent, commençant dès les petits déplacements. Dans tous les cas on observe un phénomène d’écrouissage confirmant le phénomène de butée du sol sur des inclusions longues et flexibles.

Commentaires sur les réseaux chargés latéralement.

Les essais de chargement horizontal n’ont concerné que des réseaux simples : doubles chevalets sur le site de St Rémy lès Chevreuse, simples chevalets sur les sites en Alabama (Etats Unis) et à St Maurice. Ils confirment que l’inclinaison des micropieux a un large effet bénéfique sur la résistance aux efforts latéraux. Ces cas ne sont que des chargements statiques, mais les résultats sont similaires pour des chargements dynamiques ou sismiques. En ce qui concerne les doubles chevalets de St Rémy-lèsChevreuse, la résistance horizontale est 2 à 3 fois plus grande que celle du groupe avec un espacement relatif S/B = 2. Les études numériques effectuées au CERMES ont confirmé ce résultat. 5.6

5.6.1. Groupes de micropieux Les résultats expérimentaux ont montré un effet de groupe positif (Ce > 1) pour les groupes comprenant un grand nombre de pieux flexibles, effet qui est principalement dû au confinement du sol entre les micropieux. Ce point est confirmé par les fortes valeurs du frottement latéral dans les cas où il a pu être mesuré. L’effet de groupe atteint un maximum pour un espacement relatif entre micropieux S/B = 2,5 à 4. Pour des valeurs plus élevées, le confinement est réduit et la capacité portante du groupe tend vers la somme des capacités portantes des micopieux isolés (Ce =1). Il est évident que le confinement peut être amélioré par une méthode appropriée de mise en place des micropieux (battage ou fonçage dans les sables lâches par exemple). En ce qui concerne les reprises en sous-œuvre, il est confirmé, comme l’ont montré les travaux et les études sur le Pont de Pierre à Bordeaux, que les micropieux sont une solution efficace et adaptée pour stabiliser les mouvements des fondations des structures anciennes 5.6.2 Résistance des groupes de micropieux aux efforts horizontaux. Les expériences conduites sur des groupes de micropieux chargés horizontalement montrent que les effets de groupe sont comparables à ceux de pieux de diamètres conventionnels : - la résistance totale d’un groupe de micropieux est inférieure à la somme des résistances de tous les micropieux à cause de l’effet d’ombre des pieux de devant sur les micropieux situés derrière, mais cet effet négatif peut être négligé lorsque l’espacement atteint 6 à 7 diamètres ; - lorsque les micropieux sont placés en une rangée perpendiculaire à la direction du chargement, la résistance du groupe est diminuée par les interactions mécaniques dans le sol. Cette diminution est cependant modérée et peut être négligée lorsque l’espacement dépasse 3 diamètres ; - des micropieux mis en place par refoulement du sol présentent une plus grande raideur dans un chargement horizontal que des micropieux mis en place par des techniques ne refoulant pas le sol.

Figure 9. Réseau à 18 micropieux du Laboratoire 3S de Grenoble.

5.5

347 pages Synthèse des résultats et recommandations du Projet National sur les micropieux, édité par les Presses des Ponts. Une traduction en anglais a été éditée par l’Association ADSC aux Etats Unis pour le compte de la Federal Highway Administration. Par ailleurs, l’ensemble des résultats scientifiques de l’ouvrage de synthèse a servi de base à plusieurs recherches complémentaires dans des universités étrangères. Sur un autre plan, la recherche collaborative menée à l’occasion du Projet National FOREVER a suscité la création d’une Société Internationale des Micropieux (ISM – International Society for Micropiles) regroupant les praticiens d’Amérique du Nord, d’Europe et du Japon.

Conclusions et recommandations.

Les travaux du PN FOREVER ont fait l’objet de plus de 70 rapports et articles. Ils ont abouti à la rédaction de l’ouvrage de

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5.6.3 Méthodes numériques pour estimer les déplacements d’un groupe de micropieux. Les recherches effectuées dans FOREVER ont permis le développement du programme GOUPEG qui utilise les fonctions de transfert (t-z) et (p-y) ainsi que l’élasticité linéaire pour les interactions entre les micropieux. Ce programme est bien sûr également valable pour les pieux. Pour calculer le déplacement des groupes de micropieux, il faut distinguer deux types d’effet de natures différentes : - les effets dus à la technique de mise en place qui modifie les propriétés du sol au voisinage et à l’interface sol/micropieu (effets qui ne peuvent être qu’estimés car impossibles à calculer) ; - l’effet dû à l’interaction mécanique entre les micropieux qui est ajouté aux déplacements

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5.6.4. Réseaux de micropieux En dépit du fait que les études et essais, réalisés par FOREVER ne soient pas suffisamment exhaustifs, on peut néanmoins donner les conclusions suivantes : - Un réseau, quel que soit son nombre de micropieux, a un meilleur comportement que le groupe équivalent. - En ce qui concerne le comportement sous charge verticale, les résultats expérimentaux sont pour le moins contradictoires. - Pour obtenir un effet de réseau positif, les recommandations faites pour les groupes doivent bien sûr être suivies, particulièrement en ce qui concerne le nombre et la longueur des micropieux ainsi que le confinement du sol. - Dans les sols granulaires lâches à moyennement denses, qui sont les plus avantageux à renforcer par micropieux, il est possible d’obtenir un effet de réseau positif en comparaison avec le groupe équivalent si on assure un confinement adéquat du sol et, également, si les micropieux sont concentrés autant que possible directement sous la charge appliquée. Cela implique que les micropieux ne « sortent » pas de la surface de la fondation, mais au contraire se dirigent vers l’intérieur ( < 0), pour assurer un « clouage » maximum du sol. Cela est assez similaire au concept proposé par Lizzi : une fondation de sol renforcé se comportant comme un monolithe. - Pour les sols granulaires denses qui sont difficiles à compacter, il n’est pas possible d’obtenir un effet de réseau positif. - Il n’est pas possible à l’heure actuelle de dimensionner un réseau de micropieux, sauf s’il s’agit d’un réseau simple (chevalet). Cependant des méthodes se développent actuellement utilisant les fonctions de transfert ou les techniques d’homogénéisation. - D’un point de vue pratique, l’idée qui prévalait à la fin de FOREVER était qu’il était plus avantageux de ne chercher un effet de réseau que dans le cas des micropieux forés et injectés par gravité. Pour les micropieux injectés sous forte pression du type IRS (injection répétitive et sélective), il est raisonnable de penser qu’ils travailleront plus isolément en groupe ou en en réseau simple. 5.6.5. Comportement sismique des micropieux. L’analyse des dommages causés par des séismes, comme ceux de Loma Prieta et de Kobé, a montré que les fondations qui utilisaient des pieux en acier de petit diamètre ont mieux résisté aux sollicitations sismiques que les pieux en béton de large diamètre. Cette observation plaide en faveur de l’utilisation de micropieux pour les fondations en zone sismique car ils présentent à la fois flexiblité, ductilité et résistance à la traction. Les micropieux s’avèrent particulièrement intéressants pour réparer des structures qui ont subi des dommages lors de tremblements de terre. Cette technique offre en effet aux ingénieurs beaucoup de possibilités dans le dimensionnement (nombre, inclinaison et arrangement des micropieux) ainsi qu’une facilité de mise en place qui rend son utilisation compétitive, en particulier dans les zones d’accès difficile. L’utilisation des micropieux comme technique de renforcement (groupes et réseaux) présente beaucoup d’avantages supplémentaires car elle permet de créer un composite sol/structure doté de propriétés mécaniques particulières concernant la rigidité, la résistance et avant tout la stabilité durant les tremblements de terre, en particulier dans les sites présentant un risque de liquéfaction du sol. La recherche faite par FOREVER sur ce sujet a inclu des essais en centrifugeuse, des modélisations tridimensionnelles aux éléments finis et également de simples modèles avec ressorts et dashpots (voir Shahrour et Juran, 2004). Elle a permis une meilleure compréhension du comportement des micropieux sous sollicitaion sismique. Les principaux résultats obtenus sont les suivants : a) Les efforts transmis aux micropieux résultent d’une interaction cinématique et d’une interaction inertielle.

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L’interaction cinématique est modérée pour des micropieux verticaux utilisés comme éléments de fondation. La grande flexibilité des micropieux permet le calcul des efforts dus à l’effet cinématique en supposant que les micropieux suivent le déplacement du sol en champ libre. b) Les forces d’inertie, résultant de l’accélération de la structure, transmettent au groupe de micropieux une force latérale et un moment de renversement. Les efforts latéraux et les moments de renversement provoquent des forces de compression et de traction dans les micropieux. Il est donc nécessaire de dimensionner les micropieux pour qu’ils résistent à ces forces et de prendre les mesures nécessaires pour que la fixation entre le micropieu et la semelle résiste aux forces de traction. Il faut noter que ce phénomène plaide en faveur de l’utilisation des micropieux dans les zones sismiques. c) Les systèmes de micropieux présentent un effet de groupe positif qui peut être attribué à un effet de structure résultant de la fixation des micropieux dans la semelle. Cet effet résulte de la réduction du moment de flexion dans les micropieux et des déplacements en tête lorsque l’espacement entre micropieux décroit. En l’absence de quantification, cet effet peut être négligé car il est conservatif. d) L’absence de dommages observée dans plusieurs tremblements de terre montre un comportement favorable des pieux inclinés et flexibles. Les études effectuées par FOREVER montrent que l’inclinaison des micropieux conduit à une augmentation de la raideur de la fondation par rapport au chargement sismique et à une augmentation des forces axiales dans les micropieux. e) L’utilisation de micropieux dans les sols liquéfiables présente un grand intérêt. En effet les résultats obtenus en centrifugeuse montrent que les micropieux confinent le système sol/micropieux, ce qui a pour effet de réduire le mouvement du sol, de retarder le développement de la pression interstitielle et ainsi de réduire le risque de liquéfaction. f) La comparaison des résultats des essais en centrifugeuse avec ceux de la modélisation par éléments finis et avec ceux des méthodes de calcul simplifiées basées sur le modèle de Winkler montre que ces dernières peuvent être utilisées pour le dimensionnement sismique des micropieux en fondation. g) Le dimensionnement des micropieux en zone sismique doit prendre en compte tous les autres paramètres du projet, notamment les fréquences (chargement, structures, couches de sol, etc.). 6 6.1

LE PROJET NATIONAL VIBROFONÇAGE Introduction

Le Projet National Vibrofonçage a été piloté par l’IREX à la suite d'une étude exploratoire (mars 1998), puis d'une étude de faisabilité (janvier 1999). Les conclusions du PN ont été présentées en septembre 2006. La journée de restitution était associée au symposium international TRANSVIB 2006.. Le budget global de ce projet était de 1 152 000 euros H.T., dont une subvention de la Direction de la Recherche du Ministère des Sciences et des Techniques de 246.000 euros H.T., le solde en apports en nature et cotisations des partenaires. La plus grande part de ce budget a été consacrée aux expérimentations et mesures sur sites. Faisant suite au Projet National TUBA, consacré au fonçage de pieux par battage, ce P.N. s’est intéressé à la technique plus récente de fonçage d’éléments métalliques linéaires (tubes, palplanches) dans le sol par vibrage (Figure 10).

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vibrage ont été équipés en tête de jauges de déformation et d’accéléromètres (équipement de contrôle de battage développé par TNO).Les mesures effectuées en tête seulement n’ont pas donné lieu à une interprétation détaillée. Les essais du Havre réalisés en décembre 2002.,sur un site mis à disposition par le Port Autonome du Havre dans la zone du complexe pétrochimique à proximité du pont de Normandie. Les coupes des terrains et leurs caractéristiques géotechniques sont résumées dans le tableau 2 ci-après :

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Figure 10. Vibrofonçage : Représentation schématique (Holeyman, 2002)

Il s’est conclu en 2006 par l’édition d’un Guide Technique du Vibrofonçage, simultanément en français et en anglais distribué aux participants du Symposium international « Transvib 2006 » qui s’est tenu en septembre 2006 à Paris, et a été largement diffusé depuis. 6.2

Programme réalisé

Les études et travaux réalisés au cours du P.N. Vibrofonçage ont comporté trois tranches ayant chacune plusieurs phases : - Tranche 1 : enquête sur les pratiques, synthèse des recherches antérieures et préparation de la tranche 2 incluant des essais sur site et des expérimentations ; - Tranche 2 : réalisation d’essais instrumentés de vibrofonçage et de chargement de pieux sur sites, et d’essais en chambre d’étalonnage en laboratoire ; - Tranche 3 : analyse et interprétation des résultats des expérimentations, mise au point d’un code de calcul de prévision de vibrofonçage (logiciel BRAXUUS), rédaction d’un guide technique, valorisation des résultats (organisation de Transvib 2006). TABLEAU 1 : Caractéristiques géotechniques – site de Montoir

Les expérimentations de la tranche 2 ont eu lieu sur quatre sites : Les essais de Montoir réalisés en Août 2001.Un plot d’essais grandeur nature a été réalisé à l’occasion du prolongement du Terminal à marchandises diverses et conteneurs du port de Montoir (Port Autonome de Nantes–Saint Nazaire). Les coupes des terrains et leurs caractéristiques géotechniques sont résumées dans le tableau 1 ci-après : Deux tubes métalliques fermés à la base de 339 mm de diamètre et 14 mm d’épaisseur, (longueur 32m) instrumentés à plusieurs niveaux (jauges de contraintes, accéléromètres) ont été foncés par vibrage. L’un des pieux a été surbattu pour apprécier sa portance par un essai dynamique. Un essai de chargement statique à été réalisé sur l’autre pieu, pour comparaison avec les résultats d’un essai de chargement statique réalisé sur un pieu battu de même type sur le même site en 1999. Les essais de Dunkerque réalisés en Janvier 2002.Trois tubes ouverts à la base d’un ouvrage en cours de réalisation foncés par

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Une palplanche PU16 (longueur 14m) et une sonde (longueur 14,5m) provenant d’une expérimentation antérieure (sonde SIPDIS) ont été mises en place. La sonde était instrumentée en trois niveaux, la palplanche en tête et en pied, un tube en tête et en pied, l’autre en tête seulement. Des mesures de vitesses particulaires en surface du sol ont été effectuées pendant la mise en place des deux tubes et de la sonde. Les essais de Merville réalisés de mars à juin 2003.sur le site expérimental de l’aérodrome de Merville géré par le laboratoire des Ponts et Chaussées Ils ont eu pour but de mesurer comparativement les comportements d’éléments battus et foncés par vibrage dans l’argile des Flandres. Les coupes des terrains et leurs caractéristiques géotechniques sont résumées dans le tableau 3 ci-après : TABLEAU 3 : Caractéristiques géotechniques – site de Merville

Deux tubes ouverts (longueur 12,3m) de diamètre 508mm et deux paires de palplanches AU16 (longueur 13m) ont été mis en place dans les conditions d’un chantier expérimental. Pour chaque type d’élément, l’un a été foncé par vibrage au moyen d’un vibrateur ICE 815 et l’autre battu à la même profondeur avec un marteau IHC S70.Les éléments étaient instrumentés en tête et en pied. La force de retenue, la longueur de la fiche, la pression et le débit du groupe hydraulique pour le vibrofonçage, l’énergie du marteau pour le battage, et les vitesses particulaires à la surface du sol à des distances de 5, 10 et 15 mètres de l’élément ont été mesurés en continu au cours de son enfoncement. Chacun d’eux a ensuite été soumis à un essai de chargement statique instrumenté afin de comparer la portance obtenue pour chacun des deux types de mise en place. En complément, des essais de modélisation physique du processus de fonçage par vibrage ont été menés dans la chambre d’étalonnage du CERMES au laboratoire de l’ENPC à Marne la Vallée. Une sonde prototype de fonçage par vibrage a été développée, qui peut être enfoncée dans un massif de sable reconstitué en chambre d’étalonnage grâce à un servovérin hydraulique. La sonde, d’une section droite de 10cm² (standard pénétrométrique) est instrumentée pour mesurer la résistance en pointe , ainsi que le frottement local sur un manchon spécifique. Elle est, de plus, équipée d’un accéléromètre en pointe. L’étude paramétrique réalisée par des essais à force contrôlée et à déplacement contrôlé a mis clairement en évidence l’influence des paramètres de base (force statique moyenne, amplitude et fréquence de la force cyclique) sur le déroulement du processus. Ces essais constituent un modèle physique qui peut être simulé à l’aide de logiciels et, en particulier, du logiciel BRAXUUS, développé au cours du P.N.

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Résultats remarquables

6.4.1. Pénétrations dans le sol Nous nous contenterons de souligner ici, parmi les nombreuses données expérimentales et les interprétations auxquelles elles ont donné lieu, ce qui nous parait novateur et susceptible de nous faire progresser dans la connaissance et la compréhension des phénomènes observés. En premier lieu, toutes les mesures des variables fonction du temps les ont montrées périodiques, et peuvent donc être décomposées en une valeur constante pendant la période considérée (valeur moyenne) et une fonction du temps dont la valeur moyenne sur une période est nulle. La vitesse d’enfoncement peut être supposée constante sur une période, et l’accélération moyenne nulle. Mais les mesures montrent aussi que l’effet des vibrations ne se résume pas à diminuer les frottements le long du fût du pieu : la pénétration n’est pas due au seul poids de l’ensemble pieu + vibrateur + pinces. Les variations périodiques et alternatives des vitesses particulaires du pieu servent aussi à mobiliser les forces de frottement pour aider à vaincre la résistance du sol sous la pointe du pieu, comme on peut le constater sur les graphiques représentant les valeurs moyennes des forces de frottement et de la résistance en pointe pour les essais de Merville (pieu-tube et palplanche) (Figure 11).

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En effet, l’interprétation des mesures permet de séparer l’effort exercé par le sol sous la pointe du pieu des effets latéraux sur le fût du pieu. Deux méthodes différentes ont été utilisées pour cela au cours du P.N. Nous attirons l’attention sur l’analyse exposée par Dominique Vié dans les actes du Symposium Transvib 2006 (LCPC, ISBN 2-7208-2466-6, p.195-208) .La méthode exposée, basée sur une analyse rigoureuse des vibrations enregistrées par les mesures, devrait, à notre avis, s’imposer pour l’interprétation des mesures faites sur les chantiers quand on dispose d’enregistrements en tête et en pied ou à plusieurs niveaux (dont un hors sol) d’un pieu. 300

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Guide technique 2006 : Vibrofonçage – Vibratory pile driving, Presses des Ponts, ISBN 2-85978-423-3, 282 pages. Le guide technique 2006 Vibrofonçage donne des réponses, à la lumière des connaissances actuelles et des résultats expérimentaux, aux questions que se posent les utilisateurs de la technique du vibrofonçage : - choisir la technique et le matériel ; - prévoir la fiche et les rendements (logiciel BRAXUUS) ; - apprécier les nuisances possibles des travaux ; - estimer la portance des éléments après leur mise en place. Il comprend un texte en français et en anglais décrivant les matériels, leurs choix, les logiciels de calcul existant, une analyse des essais de fonçage et de portance, une bibliographie et une liste des normes et recommandations ainsi que des rapports internes du projet national. L’annexe A rassemble des éléments théoriques. L’annexe B présente la synthèse des expérimentations en grandeur réelle et en laboratoire. L’annexe C présente les logiciels de prévision de vibrofonçage et en particulier le logiciel BRAXUUS du projet national (fourni sur un Cdrom avec le guide technique). L’annexe D présente les documentations de constructeurs partenaires du projet. Holeyman A., Vanden Berghe J.-F., Charue N. (2002) TRANSVIB 2002 : Vibratory pile driving and deep soil compaction, Balkema, ISBN 90-5809-521-5, 233 pages. Gonin H., Holeyman A., Rocher-Lacoste F. (2006) TRANSVIB 2006 : Actes du Symposium International sur le Vibrofonçage et la Vibrocompaction, publié par le LCPC, ISBN 2-7208-2466-6, 400 pages. TRANSVIB est un symposium international réunissant périodiquement toutes les personnes et organismes intéressés par le vibrofonçage des pieux et des palplanches et le compactage en profondeur des sols. Il donne lieu à la publication d’actes. Le premier a eu lieu en 2002 en Belgique à Louvain-la-neuve, le second en 2006 en France à Paris pour la valorisation et dans la continuité du projet national vibrofonçage. Il serait hautement souhaitable qu’une troisième édition ait lieu dans un avenir proche…

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A titre d’exemple, nous donnons sur les figures ci-après les graphiques Force-déplacement (composantes périodiques seulement) obtenus à Merville (Figure 12) pour le tube et une palplanche, et pour les tubes de Montoir (Figure 13), aussi bien pour la pointe que pour la résultante du frottement latéral .On notera les formes d’ellipses quasi-parfaites obtenues à Merville (Figure 3), qui peuvent être fidèlement modélisées par une loi visco-élastique linéaire, alors que la modélisation par une loi élasto-plastique est moins évidente pour les essais de Montoir. 6.4.2 Force portante des pieux vibrofoncés. Suite aux travaux bibliographiques et aux données expérimentales recueillies en réalisant des essais instrumentés avec une chaîne extensométrique, en vraie grandeur (Figure 14),

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(Luxembourg). Ces travaux ont donné et donneront lieu à publication des thèses de doctorat suivantes : - Hanus V. (2010) Analysis and modelling of the noise generation during vibratory pile driving and determination of the optimization potential, Université du Luxembourg. - Rocher-Lacoste F. (2008) Etude expérimentale en vraie grandeur et étude numérique des pieux vibrofoncés : Vibrations dans l’environnement et capacité portante, ENCP, France. - Whenham V. (2011) A study on energy transfers during pile vibratory driving, Université Catholique de Louvain & CSTC, Belgique. D’une manière prospective, on peut souhaiter, outre l’organisation d’un nouveau Transvib, la multiplication des instrumentations sur chantier et l’utilisation systématique des méthodes d’interprétation mises au point pour le P.N., et la poursuite d’un programme d’essais en chambre d’étalonnage jumelé avec l’utilisation de BRAXUUS, à priori bien adapté pour leur modélisation. Des essais grandeur nature à l’aide d’un micro pieu spécialement instrumenté seraient également très utiles sur les chantiers pour en tirer des enseignements pratiques, et cela pour un investissement raisonnablement limité dans le cadre d’une recherche appliquée.

Figure 14. Dispositif de réaction de chargement statique – Merville

des recommandations ont été déduites pour dimensionner les pieux mis en œuvre par vibrage. Par rapport aux tubes battus ouverts, aux pieux H battus et aux palplanches battues, il a lieu de faire un abattement de 30% sur le frottement latéral et de 50% sur la résistance de pointe. Ces recommandations ont été utilisées pour la rédaction de la norme d’application nationale de l’Eurocode 7 : dimensionnement et justification des fondations profondes (NFP94-262).

7 LE PROJET ASIRI SUR LE RENFORCEMENT DES SOLS DE FONDATION PAR INCLUSIONS RIGIDES.

6.4.3 Impact sur l’environnement et nuisances Le projet national a cherché à comparer les nuisances acoustiques et vibratoires entre le fonçage par vibrage et par battage pour mieux les comprendre. Une bonne connaissance des niveaux et des puissances acoustiques des matériels de fonçage par battage et par vibrage est nécessaire pour réduire cette nuisance et rendre le chantier le plus furtif possible. L'étude bibliographique et l’analyse de cinq chantiers de fonçage à permis de dire que la puissance acoustique qui caractérise les engins de chantier est supérieure de 5 à 20 dB(A) pour les moutons de battage et les trépideurs par rapport aux vibrateurs. Le fonçage par vibrage et par battage de pieux engendre des ondes dans le sol. Une étude réglementaire a été effectuée en comparant 14 règlements, d'ou il ressort une grande disparité. La bande de fréquence réglementée pour les nuisances vibratoires est comprise entre 1 et 100 Hz, et pour ce qui est des vitesses particulaires, les seuils imposés vont de 1 à 100 mm/s. Généralement, les normes nationales comportent trois bandes différentes de seuils suivant le type de structures soumises aux vibrations. En conclusion, on peut noter que, parmi les niveaux imposés par les différentes normes européennes, les seuils français sont plus sécuritaires que la moyenne des normes étudiées. Pour essayer de mieux comprendre ce phénomène de propagation des vibrations de chantier, une modélisation 2D par éléments finis a été entreprise avec le logiciel CESAR-LCPC en dynamique linéaire. Un modèle axisymétrique a été choisi. et deux paires de AU 16 foncées par vibrage ont été étudiées. Globalement la modélisation a donné des résultats satisfaisants pour de faibles enfoncements qui représentent la majorité des travaux urbains. Mais il subsiste toutefois des incertitudes et des calages qui ont nécessité l'introduction d'un coefficient d'amortissement dans le modèle numérique (formulation de Rayleigh) pour se rapprocher de la réalité. Il est à regretter l'absence de mesures expérimentales à plus de 15 m. de la source. 6.5

Prolongements et prospective

Des travaux de recherche se poursuivent dans le domaine du P.N.au Laboratoire des Ponts et Chaussées (France), au Centre Scientifique et Technique de le Construction (Belgique) et à la Faculté des Sciences, de la Technologie et de la Communication

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7.1

Objectifs et organisation

Le concept de fondation sur un sol renforcé par inclusions rigides associe des éléments verticaux très peu déformables à une semelle ou un radier par l’intermédiaire d’un matelas, appelée couche de répartition, souvent granulaire, mais sans qu’il y ait de liaison mécanique rigide entre eux. La figure 10 montre ainsi la constitution d’une fondation sur inclusions rigides.

Figure 15. Fondation sur inclusions rigides

Cette technique permet de réduire considérablement les tassements du massif de fondation sous les charges appliquées, tout en améliorant sa stabilité. Après avoir été utilisée en Scandinavie, Royaume Uni et Allemagne principalement pour des remblais (remblais fondés sur pieux) dans des zones de sol compressible), elle s’est bien développée, notamment en France, avec des applications originales à des ouvrages de grande surface comme les dallages industriels. Le champ d’application est très large depuis les ouvrages simples jusqu’aux ouvrages exceptionnels comme les fondations du pont de Rion-Antirion en Grèce. Les objectifs du Projet ASIRI ont été les suivants, grâce à des recherches expérimentales diversifiées et à des méthodes d’analyse numérique appropriées, :

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a) Combler le manque de données expérimentales de référence et appuyer le développement spécifique observé en France vers des fondations de grande surface. b) Mieux comprendre les mécanismes de transfert de charge dans le matelas de répartition placé à la base d’un remblai sur inclusions rigides ou sous une fondation étendue comme un dallage ou un radier. c) Mettre au point des méthodes de dimensionnement : en particulier disposer de modélisations numériques détaillées de référence et élaborer des méthodes simplifiées pouvant être appliquées aux ouvrages usuels. d) Élaborer un modèle global englobant le matelas et le sol renforcé, dans lequel le sol porte une partie de la charge. e) Évaluer les effets de points durs dans le cas des dallages et pouvoir évaluer les sollicitations de flexion dans ces dallages. f) Accompagner le développement de la technique en élaborant des recommandations pour la conception, l’exécution et le contrôle des travaux de renforcement par inclusions rigides. La direction du projet a comporté un président, un vice président, un directeur scientifique et un responsable du suivi par l’IREX. Le Projet ASIRI a comporté 40 partenaires répartis entre le monde de la construction et le monde universitaire. Son budget a été de 2 389 280 € dont une subvention de la DRAST de 478 000 € et le solde par les cotisations et apports en nature des partenaires. Sa durée a été de 5 années de 2005 à 2010 7.2

Programme général

Le projet ASIRI a été développé en cinq thèmes entre 2005 et 2011 : 1) Expérimentations en vraie grandeur de remblai ou de dallages sur inclusions rigides. 2) Instrumentation d’ouvrages réels réalisés dans des conditions géotechniques variées. 3) Modèles physiques en centrifugeuse ou en chambre d’étalonnage. 4) Caractérisation complète du comportement mécanique des matériaux grossiers utilisés dans les matelas de répartition des ouvrages expérimentaux ou des modèles physiques 5) Modélisations numériques de référence. En parallèle, ont été rédigées entre 2005 et 2011 des Recommandations détaillées comportant huit chapitres. Cet important programme a été le support de 9 thèses de doctorat. Il faut enfin noter que le Projet a été nécessairement centré sur les points clés de la technique et de son dimensionnement, ce qui a imposé de délaisser des points également importants comme le chargement latéral des fondations ou les sollicitations cycliques. 7.2.1

conditions en section courante d’un ouvrage, en particulier sans effet de bord. Des instrumentations très complètes ont permis de mesurer les efforts repris sur les têtes d’inclusion et entre les inclusions, ainsi que le tassement au niveau des têtes et au sommet du matelas de répartition. Des tassomètres multipoints avaient été placés sur l’épaisseur du sol compressible, ainsi que des inclinomètres sous les talus. Des transducteurs offrant une précision de l’ordre du cm avaient en plus été mis dans les plans de mesure. Enfin les nappes de renforcement utilisées sous les remblais avaient été instrumentées par des fibres optiques. 7.2.1.2 Principaux enseignements Les deux expérimentations en vraie grandeur ont apporté une bonne amélioration des connaissances sur le comportement et le mécanisme de la technique des inclusions rigides. Parmi les points généraux ou plus particuliers, on peut faire les remarques suivantes : a) La réduction significative des tassements des ouvrages sur inclusions rigides par rapport au cas du sol non renforcé (facteur de 5 à 6) est confirmée. b) Entre les têtes des inclusions, la déformée du sol s’avère plane et il se confirme que l’efficacité en tassement est toujours meilleure que l’efficacité en contraintes. c) A la base d’un remblai fondé sur un massif de sol renforcé par inclusions rigides, une couche de répartition ou matelas de bonne qualité joue un rôle déterminant pour un bon transfert de charge entre le remblai et les inclusions. e) Une géogrille de renforcement dans la couche de répartition apporte une meilleure efficacité qu’une nappe en géotextile. Les déformations subies lors de la mise en place et du compactage de cette couche paraissent avoir un rôle déterminant (bien souligné par les fibres optiques). Un matelas de répartition renforcé par deux géogrilles s’est révélé avoir pratiquement le comportement d’une dalle « armée » reposant sur les têtes d’inclusions.

.Ouvrages expérimentaux

7.2.1.1. Spécificités Deux sites l’un à Saint-Ouen-l’Aumône, l’autre à Chelles ont permis de réaliser deux expérimentations en vraie grandeur d’ouvrages sur inclusions rigides: un remblai d’une part, des dallages supportant une charge répartie d’autre part. Chacun des deux ouvrages a comporté un plot non renforcé pour servir de référence et a fait l’objet d’essais de chargement sur des inclusions isolées. Cela a permis de juger de l’efficacité de la technique en contrainte et également en tassement. Il a par ailleurs été comparé les comportements avec des inclusions mises en place par refoulement du sol et sans refoulement. Des reconnaissances géotechniques spécifiques ont été réalisées avec des sondages carottés, des essais en place et des essais de laboratoire. Le matériau du matelas (grave industrielle) a notamment fait l’objet d’essais triaxiaux en diamètre de 300 mm qui ont permis de dresser une base de données de référence pour des matériaux graveleux. Chaque plot renforcé comportait 16 inclusions permettant d’avoir une maille centrale parfaitement représentative des

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Figure 16. St Ouen-l’Aumône.(Plot de dallage). Comparaison entre le chargement du plot expérimental et un essai de chargement d’un micropieu avec mesures en tête et en pointe

f) Le comportement observé sur la tête d’une inclusion en maille courante d’un plot duplique le comportement observé en pointe d’une inclusion isolée chargée axialement en tête, comme le montre la figure 16, ce qui est un résultat important montrant que globalement les frottements latéraux positif et négatif s’équilibrent. Mais on doit considérer qu’il n’est valable que si les inclusions reposent sur un substratum résistant. Il est donc important de pouvoir modéliser de manière correcte le comportement de la pointe d’une inclusion pour assurer une bonne représentation du modèle numérique complet. Ce résultat a amené à imposer le calage préalable des modèles numériques en simulant, dans le modèle préparé, le comportement d’une inclusion isolée sous chargement axial pour comparer la réponse

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

obtenue soit aux résultats d’un essai spécifique, soit à ceux d’une simulation semi-empirique par les courbes de transfert, dont la représentativité est bien démontrée. Ainsi est souligné l’intérêt des essais de chargement d’inclusions isolées pour une bonne conception des ouvrages. g) Les résultats de ces expérimentations montrent que le sol renforcé subit également des déformations latérales en périphérie, lesquelles doivent être prises en compte dans le dimensionnement des inclusions (nécessité d’armer ou non les inclusions placées en rive). Il a été mesuré un rapport de 0,25 entre le déplacement horizontal maximal et le tassement au centre de la zone renforcée, rapport comparable à celui applicable sous les talus des remblais sur sols compressibles. h) Ces expérimentations ont également montré l’importance d’une bonne caractérisation géotechnique des sites. Les mérites du pénétromètre statique ont été reconnus et la réalisation d’essais oedométriques est indispensable. Les essais pressiométriques permettent quant à eux une bonne corrélation avec l’expérience relative aux fondations profondes (valeurs limites du frottement latéral et/ou de la charge en pointe, allure des courbes de transfert et élaboration des courbes de chargement d’inclusions isolées). 7.2.2 Instrumentation d’ouvrages réels Les résultats des expérimentations en vraie grandeur ont été complétés par des instrumentations sur des chantiers d’ouvrages réels afin d’y collecter des données supplémentaires sur le comportement des inclusions dans des conditions variées. Plus d’une dizaine d’ouvrages ont ainsi été instrumentés parmi lesquels on peut citer : une fondation d’éolienne, un radier pour infrastructure de déchets faiblement radioactifs, un cadre en béton armé enterré, un réservoir de traitement d’eaux usées, un dallage industriel pour examiner l’incidence de charges ponctuelles (pieds de racks ou roues de chariot). Il faut signaler les difficultés inhérentes à ces chantiers, dont la plus importante est de préserver les capteurs et leurs connexions durant les phases successives de travaux. 7.2.3

Modèles physiques

7.2.3.1. Spécificités Des modèles physiques ont été mis au point en chambre d’étalonnage pour étudier le transfert de charge autour d’une tête d’inclusion, l’influence de l’épaisseur de la couche de répartition et, pour une même épaisseur de cette couche, les différences entre un dallage et un remblai pour des conditions de matelas comparable. Les modèles physiques les plus intéressants ont été ceux faits en centrifugeuse où toutes les conditions de similitude sont respectées. La capacité de la centrifugeuse de l’IFSTTAR à Nantes atteint 100g et il a été fait le choix d’un modèle au 1/28 pour étudier un groupe de 9 inclusions et d’un modèle au 1/12 pour des essais avec plateau mobile permettant de simuler le tassement du sol sur des groupes d’inclusions. En tout, 35 essais en centrifugeuse ont été réalisés pour une étude paramétrique détaillée selon le type d’ouvrage supporté (remblai ou dallage), l’espacement des inclusions, la hauteur de la couche de répartition et le type de matériau de cette couche (grave naturelle ou limon traité). 7.2.3.2. Enseignements Les modèles en chambre d’étalonnage montrent une certaine différence entre remblais et dallages pour de faibles épaisseurs da la couche de répartition, différence qui s’estompe lorsque l’épaisseur augmente. Ils montrent également que la granularité du matelas est un facteur clé. Il a également été mis en évidence une réversibilité moindre pour un remblai que pour un dallage, ce qui souligne le rôle important joué par le dallage (comportement élastique réversible) par rapport au remblai où

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les cisaillements qui accompagnent le transfert de charge sont irréversibles. Les essais de plateau mobile ont permis de valider que le modèle de Prandtl développé pour une semelle filante pouvait également être utilisé pour évaluer la contrainte limite sur une tête d’inclusion, sous un dallage. Il a également été établi que l’importance des déformations justifiait l’adoption de l’angle de frottement à l’état critique plutôt que de l’angle de frottement de pic. Ces résultats ont guidé le choix des règles de vérification explicitées dans les Recommandations et également celui des règles pour la vérification des conditions de cohérence des modèles simplifiés de dimensionnement. 7.2.4 Modèles numériques Les modèles numériques sont un accompagnement indispensable des expérimentations sur ouvrages en vraie grandeur ou en modèle réduit. Dans ASIRI ils ont compris des modèles numériques 3D aux éléments finis et aux différences finies qui doivent servir de référence. Mais un point important concerne le choix des lois de comportement et le calage des paramètres tirés des caractérisations détaillées effectués sur les différents matériaux (matelas de répartition, sol compressible) qu’il convient de mettre dans ces modèles. Ils ont servi, dans le cas des ouvrages testés, à vérifier leur capacité à reproduire correctement le comportement de ces ouvrages. Mais certains d’entre eux se sont révélés très exigeants en temps de calcul (plusieurs semaines). Ils ont également permis la vérification des effets de bord : comparaison des modèles 3D complet aux modèles 3D vrai ou 2D axisymétrique d’une cellule élémentaire. L’étude des conditions de représentativité des modèles vis-àvis de la simulation du comportement sous la pointe des inclusions (extension des modèles et nombre minimum d’éléments) a été faite. Elle montre qu’il faut choisir un compromis entre précision et durée des calculs. Les modèles ainsi évalués ont pu être appliqués à des situations d’ouvrages autres que celles des ouvrages expérimentaux. Il en a été ainsi par exemple du cas des dallages soumis à des charges en bandes ou des charges ponctuelles (pieds de rack) et aussi du cas des semelles situées sur un nombre limité d’inclusions et soumises à des chargements quelconques (cas non traité expérimentalement mais qu’il était indispensable d’étudier pour en donner les résultats dans les Recommandations, car ces ouvrages sont courants dans les projets d’entrepôt industriel ou logistique). Le projet ASIRI a également développé des modèles en éléments discret. Il est intéressant d’indiquer qu’ils ont révélé une meilleure aptitude que les modèles continus à décrire le comportement du matelas de répartition observé dans les modèles physiques (glissement des particules au bord des têtes d’inclusion). Mais leur mise en œuvre reste lourde et doit être réservée à des études particulières d’étalonnage ou de validation. 7.3

Publications du Projet ASIRI.

Les travaux du Projet ASIRI ont fait objet de nombreux rapports internes qui ont été présentés à l’occasion de plus de 20 conférences nationales et internationales. Ils ont par ailleurs donné lieu à 9 thèses de doctorat. Un livre très documenté intitulé Recommandations pour la conception, le dimensionnement, l’exécution et le contrôle de l’amélioration des sols de fondation par inclusions rigides a été publié par les Presses des Ponts en 2012. Il comprend 384 pages et huit chapitres :1. Description et développement jusqu’au projet national – 2. Mécanismes et fonctionnement – 3. Modèles de calcul – 4. Conception – 5. Justifications – 6. Reconnaissance des sols – 7. Exécution – 8. Contrôles et instrumentations.

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7.4

Les retombées du Projet ASIRI.

Le projet ASIRI a mis en lumière comment un matelas de répartition granulaire coiffant un réseau d’inclusions rigides opère un report de charge pérenne et significatif. L’application à des dallages connaît actuellement un grand développement et constitue une spécificité au plan international. L’intérêt pour cette technique a favorisé son utilisation pour des ouvrages sensibles. Il convient de citer le projet ICEDA d’entrepose de déchets radioactifs, soumis aux exigences des installations nucléaires, qui a pu émerger et être concrétisé, après aval des autorités de sûreté, en partie grâce au capital d’expérience accumulé sur ce type d’ouvrages par le Projet ASIRI. 8

LE PROJET NATIONAL SOLCYP

SOLCYP est l’anagramme de Pieux sous SOLlicitations CYcliques. 8.1

Présentation générale du projet Solcyp

8.1.1. Objectifs Le projet SOLCYP vise à améliorer les connaissances sur le comportement des pieux de fondations soumis à des sollicitations cycliques. Il s’est fixé comme objectif le développement de procédures permettant la prise en compte de l’effet des cycles dans le dimensionnement des ouvrages de génie civil ou maritime. La phase ultime du projet consistera en un travail prénormatif en vue de l’introduction dans les réglementations nationales et internationales de la procédure proposée et des méthodes de calcul associées. Le projet couvre les aspects suivants: pieux battus et pieux forés ; sables et argiles ; charges verticales ou horizontales ; chargements cycliques répétés ou alternés ; grand nombre de cycles. 8.1.2. Carence règlementaire Alors que l’industrie pétrolière a développé des procédures pour prendre en compte l’effet des fortes charges cycliques dues à la houle sur les fondations des structures offshore, l’effet des sollicitations cycliques sur le comportement des fondations est largement ignoré dans le champ d’activité de la construction et du génie civil. Il existe certes quelques exceptions notoires comme l’étude de la liquéfaction des sols sous sollicitations sismiques, ou la fatigue des chaussées et des remblais ferroviaires. Mais d’une manière générale, il n’y a pas - sur le plan national, européen (Eurocodes) ou international (ISO) - de document spécifique traitant explicitement des risques liés aux sollicitations cycliques et proposant une approche méthodologique pour les prendre en compte dans le dimensionnement des fondations. 8.1.3. Ouvrages concernés Cette carence est d’autant plus surprenante qu’il existe une large gamme d’ouvrages soumis à des charges éminemment répétitives et présentant un certain degré de régularité en amplitude et période de retour. Les charges « cycliques » sont essentiellement d’origine environnementale (houle, vent, courant, marée) ou opérationnelle. On peut citer notamment: les éoliennes terrestres ; les ouvrages côtiers ou portuaires (jetées, digues, ..) ; les structures supports légères ou élancées sujettes à l’action du vent telles que : pylônes de transport d’énergie, cheminées et tours de grande hauteur ; les ouvrages d’art supportant les infrastructures de transport, notamment les ponts ferroviaires ; les fondations de grues, ponts roulants, turbines hydrauliques. Les ancrages des nouvelles structures liées au marché émergeant des énergies nouvelles (éoliennes terrestres et offshore, hydroliennes, panneaux photovoltaïques de grande

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surface) sont particulièrement sensibles à la répétitivité et au cumul des chargements. 8.1.4. Le programme SOLCYP Le programme du projet SOLCYP a été mis au point dans le cadre des travaux d’un groupe de travail émanant du Pôle de Compétence « Sols » de l’IREX. Il s’est concrétisé par la définition de deux volets complémentaires : un projet - dit ANRSOLCYP - qui a obtenu un financement auprès de l’Agence Nationale de la Recherche et un projet - dit PN-SOLCYP – organisé sous forme de Projet National avec le soutien financier du MEDDE (Ministère de l’Ecologie, du Développement Durable et de l’Energie), de la FNTP (Fédération Nationale des Travaux Publics) et de 14 maîtres d’ouvrage ou entreprises appartenant aux secteurs du génie civil et de l’énergie. Le budget total HT était voisin de 4,5 M€ se répartissant en 2.6M€ pour la partie ANR-SOLCYP et 1.9M€ pour la partie PNSOLCYP. La part de financement public s’élève à 28%. Le solde est couvert par les cotisations des partenaires et les apports en nature. Le projet a démarré au second semestre 2008 et est prévu de se terminer en 2014. L’organisation en deux volets a favorisé une forte implication d’organismes universitaires et de laboratoires de recherche publics (6 participants). Le volet ANR concernait la partie « académique » du projet comprenant : l’étude du comportement cyclique des sols de référence (argiles et sables) à l’aide d’essais de laboratoire (triaxial cyclique, DSS cyclique) ; l’étude du comportement statique et cyclique des interfaces à partir d’essais spéciaux ; la réalisation d’essais instrumentés sur modèles réduits en grande chambre de calibration et en centrifugeuse ; le développement de modèles numériques. Le volet PN est plus particulièrement consacré aux études expérimentales sur ouvrages en vraie grandeur: instrumentations d’ouvrages sur pieux, essais de pieux sur sites expérimentaux, développement d’essais in situ pour la mesure des paramètres cycliques des sols. 8.2

Apports du programme SOLCYP

8.2.1. Caractérisation des charges cycliques Dans le domaine de la construction et du génie civil, on a coutume de considérer que les charges appliquées sont de nature statique ou quasi-statique. En accord avec les textes règlementaires, les charges critiques sont définies par la valeur maximale attendue sous les différents cas de charge considérés (de service-ELS ; environnemental extrême-ELU ; accidentelELA). La réponse d’un sol sous sollicitations cycliques est complexe et dépend de plusieurs paramètres: contrainte moyenne, amplitude de la contrainte cyclique, fréquence de sollicitation, vitesse de chargement et nombre de cycles. Ces aspects sont familiers en géotechnique pétrolière offshore mais la nécessité d’une caractérisation complète et précise des chargements appliqués n’est pas bien perçue dans le domaine du génie civil au sens large. La collection de cas de charges réels et l’instrumentation de structures permettent de mieux cerner ces différents aspects. Les connaissances sur la réponse des sols aux chargements cycliques sont centrées autour des phénomènes liés aux séismes ou à la houle, c'est-à-dire mettant en jeu des nombres de cycles relativement faibles (de l’ordre de quelques dizaines à quelques milliers) et des périodes inférieures à 100 secondes. Il existe clairement un besoin d’étendre les investigations vers les grands nombres de cycles (au-delà du million pour les éoliennes et les sollicitations de trafic intense) et de pouvoir prendre en considération les phénomènes liés aux grandes périodes de retour (par exemple vis-à-vis de l’effet des marées ou des cycles de chargement/déchargement de grands réservoirs).

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Proceedings of the 18 International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

8.2.2. La base expérimentale du projet SOLCYP Des essais de pieux ont été conduits sur deux sites expérimentaux du Nord de la France. Le premier est à MERVILLE, où l’argile raide très fortement surconsolidée des Flandres est présente dès 3m de profondeur. Dix pieux d’essai ont été installés incluant quatre pieux métalliques tubulaires fermés battus, quatre pieux forés de type CFA, et deux pieux vissés. Les pieux ont 13m de fiche et des diamètres de 406mm (pieux battus) ou 420mm (pieux forés). Ils ont été soumis à des essais de chargement statique conventionnel par paliers, à des essais de chargement monotone rapide et à des séries d’essais cycliques incluant des essais à forte amplitude amenés à la rupture sous un petit nombre de cycles et des essais à faible amplitude conduits jusqu’à 10 000 cycles. Tous les modes de chargement ont été appliqués (tension, compression, répétés, alternés). Les résultats principaux ont été publiés dans Benzaria et al. 2012 et 2013a. Le second site est à LOON-PLAGE près de Dunkerque dans des sables denses. On y a installé deux pieux battus et cinq pieux forés CFA de mêmes caractéristiques qu’à MERVILLE mais de fiches différentes (10.5m pour les pieux battus et 8m pour les pieux forés). Le programme de chargement était similaire (Benzaria et al. 2013b) De nombreuses séries d’essais sur pieux modèles instrumentés ont été réalisées dans du sable de Fontainebleau et dans de l’argile Speswhite. Ces essais ont été effectués dans la centrifugeuse de l’IFSTTAR (ex LCPC) à Nantes. L’objectif poursuivi était d’établir des diagrammes de stabilité cyclique dans les deux types de sols de référence (sables et argiles) et pour les deux types de pieux considérés (battus et forés) en balayant une large gamme de conditions initiales (densité, consistance, consolidation) et de modes de chargement. Les premiers résultats ont été publiés (Guefresh et al. 2012 ; Puech et al. 2013). Ce type d’approche a permis de confirmer la représentativité des données acquises lors des essais in situ et d’en étendre la validité. Une troisième approche expérimentale a consisté à effectuer des essais sur pieux modèles très fortement instrumentés dans la grande chambre de calibration du laboratoire 3S-R à Grenoble. Ces essais en sable de Fontainebleau, conduits en collaboration avec l’Imperial College de Londres, ont fourni de remarquables informations sur la mobilisation du frottement à l’interface solpieu et son évolution avec l’intensité et le nombre de cycles (e.g. Tsuha et al. 2012 ; Silva et al., 2013). 8.2.3. Réponses des pieux aux chargements cycliques La figure 17 illustre le type de comportement observé sur le site d’argile surconsolidée de MERVILLE. Elle représente la relation charge-déplacement de la tête du pieu foré F2 sollicité en compression. La charge ultime en compression Quc mesurée par un essai statique conventionnel sur le pieu F1, identique au pieu

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F2, est égale à 900kN. Le pieu F2 – qui n’a subi aucune sollicitation préalable – est d’abord soumis à une série de trois chargements cycliques de plus de 3000 cycles. Les deux premières séries (CC1 et CC2) ne provoquent pas de déplacement permanent significatif du pieu. La troisième série (CC3) caractérisée par un chargement maximal Qmax de l’ordre de 800kN génère des déplacements permanents non négligeables (près de 20mm). L’essai est arrêté après 3000 cycles et suivi d’un chargement statique rapide (CR1) qui indique une capacité post-cyclique de 900kN. Sept séries de cycles sont ensuite appliquées. Les séries CC4 à CC7 ne provoquent pas de déplacement permanent de la tête du pieu au bout de 1000 cycles par série. (On notera que les essais ont été arbitrairement séparés pour permettre de les visualiser). Les essais CC8 à CC10, qui atteignent à nouveau un effort maximal de 800kN génèrent des déplacements permanents qui se cumulent rapidement (chaque série a moins de 100 cycles). La capacité post cyclique est toujours de l’ordre de 900kN (essais CR2 à CR4). 1 00 0

E f fo r t e n t ê te d u p ie u (k N )

Les histoires de chargements cycliques - calculées ou mesurées - appliquées aux fondations se composent d’une succession de charges variables d’amplitude irrégulière et de distribution relativement aléatoire. Cependant, les essais cycliques qu’il est possible de réaliser en laboratoire sur des échantillons de matériaux sont normalement conçus en séries de cycles d’amplitude régulière et de période constante. Un logiciel nommé « Cascade » permettant de transformer une série aléatoire de charges cycliques en une succession de séries ordonnées d’amplitudes constantes a été développé dans le cadre du projet. Il est basé sur l’utilisation des méthodes de comptage de cycles, de type rainflow ou “en cascade” (ASTM E 1049-85, NF A03-406, 1993). Le concept de dommage au sens de Miner est alors appliqué pour estimer l’endommagement du matériau à partir de courbes de typ S-N (dites aussi courbes de Wöelher) obtenues expérimentalement en amenant à la rupture des échantillons soumis à des séries de cycles d’amplitude de contrainte constante.

CR1

90 0 C C1 à 3

CR2 CC4

80 0

à CC1 0

70 0 60 0 50 0 40 0 30 0 20 0 10 0 0 0

50

1 00

1 50

D é p l ac e m e n t e n t ê te d u p i e u ( m m )

2 00

CR : essais monotones rapides ; CC : essais cycliques répétés CC1, 2 : N>3000 ; CC3 : N=3000 ; CC4 à 7: N=1000 ; CC8 à 10 : N1000) ; - dès que ce seuil est atteint, des déplacements permanents sont générés et la rupture cyclique intervient rapidement, typiquement en moins de 100 cycles ; - le seuil est élevé dans le domaine des chargements répétés (80 à 90% de Qus) mais décroit dans le domaine des chargements alternés (voir figure 3) ; - la capacité post cyclique n’est pas affectée par les chargements cycliques préalables. Le comportement observé sur le site de sable dense de LOON-PLAGE est très différent. La figure 18 montre la réponse de deux pieux forés identiques F4 et F5. Le pieu F4 a été soumis à un essai de chargement statique conventionnel qui indique une charge ultime de référence Quc= 1100kN. Le pieu F5 a été soumis à un chargement cyclique caractérisé par une valeur Qmax ~ 0.62 Quc. Le pieu cumule très rapidement des déplacements permanents (3% de déplacement relatif après seulement 14 cycles). L’essai a été stoppé et l’amplitude cyclique fortement réduite (Qmax ~ 0.35 Quc). Le pieu a continué à cumuler des déplacements (14mm en 5000 cycles). D’une manière générale on a observé que :

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les pieux forés étaient très sensibles aux chargements cycliques ; la capacité post cyclique était fortement affectée par les chargements cycliques ; les critères de rupture cyclique en compression devaient être définis en termes de déplacements tolérables. L’analyse de l’ensemble des essais a conduit à définir la rupture cyclique pour un déplacement relatif de 3%.

-

Effo rt e n tête d u pie u (kN )

9 00

F4 S t a t ique

8 00

Q u c = 1 1 0 0k N à 4 2 mm

F5 -C C1 N= 14

7 00 6 00 5 00

F5 -C C2 N=5 0 0 0

4 00

Figure 19. Diagramme de stabilité cyclique pour un pieu foré dans l’argile surconsolidée des Flandres à MERVILLE (Benzaria 2012)

3 00 2 00 1 00 0 0

5

10

15

20

25

D ép la cem e nt en tête d u pi eu (m m )

30

Figure 18 Relations effort-déplacement de la tête lors d’essais répétés en compression sur le pieu foré F5 de MERVILLE (d’après Benzaria et al. 2013b). Comparaison avec l’essai statique conventionnel du pieu F4.

8.2.4. Diagrammes de stabilité cyclique Le résultat d’un essai cyclique de pieu peut être avantageusement synthétisé dans un diagramme d’interaction cyclique. Chaque essai y est représenté par un point dans un diagramme Qcy/Qus en fonction de Qa/Qus avec Qcy = demiamplitude cyclique ; Qa (= Qm) = charge moyenne ; Qus = charge statique ultime de référence. Qus est déterminée à partir d’un essai statique conventionnel réalisé en compression (Quc) ou en tension (Qut) selon le mode de chargement considéré. Chaque point est affecté du nombre de cycles (Nf) ayant conduit le pieu à la rupture selon un critère donné ou du nombre de cycles total appliqué au pieu. Lorsque l’on dispose d’un nombre suffisant d’essais (de l’ordre de 10 à 20) on peut délimiter des zones dans lesquelles la stabilité du pieu est assurée pour un certain nombre de cycles et des zones dans lesquelles la rupture cyclique est obtenue pour un nombre de cycles réduit. On peut alors parler de diagramme de stabilité cyclique. Un apport essentiel du projet SOLCYP est l’établissement de diagrammes de stabilité cyclique pour différents types de pieux : battus, forés CFA, vissés ; des argiles surconsolidées et des sables denses ; des chargements répétés (en compression et en tension) et des chargements alternés ; des nombres de cycles importants (jusqu’à 10 000 par série). Ces diagrammes ont été obtenus à partir des essais in situ sur les pieux expérimentaux mais également à partir des essais sur pieux modèles. On donne à titre d’illustration deux diagrammes obtenus l’un sur des pieux forés dans l’argile des Flandres et l’autre dans les sables denses de Dunkerque. Dans l’argile des Flandres, comme indiqué plus haut, la transition entre zone stable et zone instable est brutale. Elle se traduit par une ligne unique bien définie dans le diagramme de stabilité (Figure 19). Cette ligne traduit la dépendance du seuil de chargement critique avec la valeur de la charge moyenne. On vérifie bien que dans le domaine des essais répétés le seuil se situe à des valeurs de Qmax = Qa+Qcy élevées (Qmax/Quc> 0.8). En mode alterné, la zone d’instabilité n’a pu être explorée de sorte que la ligne en pointillé apparaît comme une enveloppe conservative de la zone de stabilité. La détermination de diagrammes de stabilité dans les sables a été faite en prenant comme charge statique de référence la capacité statique mesurée juste avant la séquence cyclique considérée et en adoptant un critère de rupture cyclique du pieu égal à 3% de déplacement relatif en tête.

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Figure 20. Diagramme de stabilité cyclique pour un pieu foré dans le sable dense de Dunkerque (Loon-Plage). Puech et al. 2013.

La figure 20 montre le diagramme de stabilité cyclique des pieux forés à LOON-PLAGE (Puech et al. 2013). On est amené à définir trois zones : - une zone instable dans laquelle les pieux évoluent rapidement vers la rupture (Nf < 100), - une zone stable correspondant à des chargements cycliques de faible amplitude dans laquelle les pieux peuvent supporter plus de 1000 cycles sans accumuler de déplacements permanents significatifs, - une zone intermédiaire dite métastable dans laquelle le pieu cumule des déplacements importants ou parvient à la rupture cyclique entre 100 et 1000 cycles. La plus grande sensibilité aux chargements cycliques dans les sables est clairement visible. Il est intéressant de noter qu’une série d’essais sur pieux modèles en centrifugeuse dans lesquelles huit pieux moulés étaient mis en place dans un même conteneur de sable dense de Fontainebleau et sollicités de manière unique (un seul essai statique ou cyclique par pieu) a conduit à un diagramme de stabilité quasiment identique à celui des pieux forés à LOON-PLAGE. L’ensemble des résultats SOLCYP montre que la réponse cyclique des pieux dépend plus ou moins fortement du type de pieu, des conditions de sol, du mode de chargement et de l’histoire des chargements. 8.2.5.

Méthodologie de dimensionnement sous chargement cyclique axial Le dimensionnement complet d'un pieu soumis à des chargements cycliques axiaux peut faire appel à des procédures relativement complexes qui ne sont pas toujours nécessairement justifiées par la pratique quotidienne. SOLCYP a pris le parti de proposer différentes approches correspondant à des niveaux de complexité croissante et de développer des critères pour

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'

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 =27

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P(N)/P(1) = 1+t.Ln(N)

P(N) et P(1) sont les valeurs de la propriété décrite aux cycles N et 1 respectivement ; m et t sont des fonctions des caractéristiques du chargement (Qa et Qcy), de la rigidité du système sol-pieu et du mode d’installation.

La figure 22 illustre le processus de calage d’une loi d’évolution du déplacement de la tête d’un pieu y en fonction du nombre de cycles sous la forme y(N)/y(1) = f(N) pour deux essais de pieux-modèles en centrifugeuse dans une argile normalement consolidée. Un travail de ce type a pu être effectué dans les sables et les argiles tant pour les déplacements que pour les moments (Khemakhem et al. 2012 ; Rosquoët et al. 2013). Une synthèse est présentée dans Garnier, 2013.

0

0

P(N)/P(1) = k.Nm

Figure 22: Essais en centrifugeuse ; argile normalement consolidée. Comparaison de courbes calculées et expérimentales pour les déplacements normalisés yn/y1 de la tête de pieu. Khemakhem et al. 2012.

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L’objectif essentiel de ces essais était de dériver des lois de dégradation pour des analyses de type global ou pour la construction de courbes p-y dégradées. L’effet des cycles se traduit essentiellement par un cumul de déplacements de la tête du pieu et par un accroissement progressif du moment maximal. Les méthodes dites globales consistent à décrire l’évolution de ces phénomènes par des lois du type :

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Contrainte radiale r (kPa)

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Figure 21: Chemins de contraintes sur la paroi d’un pieu modèle ; sable de Fontainebleau dense ; essai métastable, 1000 cycles ; Silva et al (2013)

Ces chutes de contraintes ont pu être simulées en laboratoire (Pra-ai, 2013) par des essais cycliques à rigidité normale imposée (essais CNS). Les efforts portent actuellement sur la formulation de la base de données sous la forme de lois d’interface pouvant être introduites de manière « explicite » dans des modélisations par éléments finis. 8.2.6. Pieux sous chargement cyclique latéral La réponse des pieux sous chargement cyclique horizontal n’a été étudiée expérimentalement dans le cadre de SOLCYP que dans le cas de pieux flexibles et en centrifugeuse. De nombreuses séries d’essais répétés et alternés ont été effectuées sur des pieux modèles instrumentés moulés en place dans du sable de Fontainebleau (Rakotonindriana, 2009) et de la kaolinite Speswhite normalement consolidée et surconsolidée (Khemakhem et al., 2012).

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max

Réaction du sol (kN/m)

Contrainte de cisaillement rz (kPa)

déterminer quel niveau d'analyse était le plus approprié selon le cas considéré. Le diagramme de stabilité cyclique tel que décrit précédemment est un outil particulièrement intéressant pour identifier les cas de chargements critiques justifiant une analyse spécifique et plus ou moins approfondie (Jardine et al. 2012). Ce concept s’applique bien aux pieux courts et rigides, tels qu’utilisés en construction et génie civil. Dès lors, trois types d'approches peuvent être mis à la disposition du concepteur: – des approches « globales », qui ne s'intéressent qu'au comportement "global" du pieu : évaluation de l'accumulation des déplacements permanents de la tête du pieu sous l'effet des cycles ; dégradation de la capacité portante ; – des approches « locales » dans lesquelles la relation entre la contrainte de cisaillement mobilisable à l'interface sol-pieu et le déplacement local du pieu s'exprime au moyen d'une courbe de transfert dite courbe "t-z". Le défi est ici de proposer des courbes "t-z" cycliques en complément des courbes t-z recommandées par les codes actuels ; – des approches par la méthode des éléments finis. La base expérimentale de SOLCYP permet de travailler sur ces trois axes, et notamment sur la calibration des algorithmes de génération de courbes t-z proposés dans des programmes tels que RATZ (Randolph, 1994) ou SCARP (Poulos, 1989). Les essais sur le sable de Fontainebleau effectués en chambre de calibration au 3S-R ont permis de mettre en évidence que la perte de frottement sous charges cycliques était le résultat d’une chute des contraintes effectives radiales sur le pieu. Ce résultat est illustré sur la Figure 21 qui montre l’évolution en cours de cyclage des chemins de contraintes effectives (radiales et tangentielles) mesurées à trois niveaux le long de la paroi du pieu.

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Courbe P-y cyclique Courbe P-y enveloppe 0.1

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Figure 23. Relations P-y expérimentales obtenues lors d’un essai cyclique alterné (Hc/Hmax = 0,57) ; argile normalement consolidée ; Khemakhem, 2012.

Les méthodes globales sont susceptibles de fournir une réponse suffisante au concepteur dans le cas de sols homogènes et de chargements cycliques modérés. Pour des cas plus

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complexes, le recours à la méthode locale, basée sur l’utilisation de courbes de transfert dites p-y, est nécessaire. La Figure 23 illustre le phénomène de dégradation due aux cycles de la réaction locale sous la charge maximale Hmax. Cette dégradation peut être approchée par des courbes enveloppes telles que celle proposées par l’API RP2GEO avec cette double limitation que la dégradation est forfaitaire et ne dépend ni des caractéristiques réelles de la charge, ni du nombre de cycles. Rakotonindriana (2009) a montré que lorsque l’on dispose d’un nombre de données suffisantes (plusieurs essais cycliques à différents niveaux de chargement), on peut définir pour chaque profondeur un réseau de courbes P-y correspondant à un nombre de cycles donné. Ces courbes qui peuvent être réellement qualifiées de « P-y cycliques » représentent la dégradation des courbes P-y statiques qu’il faut supposer pour retrouver le comportement global du pieu après N cycles. À partir de l’importante base de données d’essais en centrifugeuse réalisés à IFSTTAR, antérieurement et dans le cadre du projet SOLCYP, on a pu montrer que la quantification de l’influence des cycles sur la « dégradation » de la courbe P-y pouvait être introduite à l’aide d'un coefficient d’abattement rc qui dépendait du nombre de cycles N, de la charge maximale appliquée Hmax et de l’amplitude des cycles Hc. Des expressions de rc ont été développées dans le cas des argiles et des sables (Khemakhem, 2012 ; Rosquoët, 2013 ; Garnier, 2013) L’approche proposée par SOLCYP constitue une avancée décisive pour la prise en compte de l’effet des chargements cycliques sur le comportement des pieux sous efforts latéraux. 8.3

Conclusion

Le projet SOLCYP s’est fixé comme objectif d’apporter une meilleure compréhension du comportement des pieux sous chargements cycliques et de développer des méthodes innovantes pour leur dimensionnement sous charges axiales et latérales. La réponse de pieux forés et battus dans les sables et les argiles a été étudiée par différentes approches tant expérimentales (en laboratoire, sur modèles et in situ) que théoriques. La plupart des données expérimentales sont actuellement disponibles mais le travail d’interprétation des données et les développements méthodologiques restent à approfondir. Pour le dimensionnement des pieux sous charges cycliques axiales, une approche graduelle est proposée. La première étape destinée à évaluer la nécessité ou non de procéder à une étude cyclique détaillée est basée sur une comparaison entre les caractéristiques des charges cycliques et le diagramme de stabilité du pieu. Un apport décisif du projet SOLCYP est de proposer des diagrammes de stabilité pour des pieux forés et battus, différents types de sols (sables et argiles) et tous modes de chargement (en compression, en tension, alterné). L’analyse complète des résultats expérimentaux de grande qualité obtenus permettra à terme de mettre à disposition du projeteur trois types d’approches : approche globale, approche locale par courbes cycliques de transfert et approche numérique par éléments finis. Pour le dimensionnement sous charges cycliques latérales, une importante banque de données en centrifugeuse a été réalisée sur sables et argiles. Des formulations ont pu être proposées pour décrire, de manière globale, la dégradation du déplacement de la tête du pieu et des moments fléchissants maximaux en fonction du nombre de cycles et de leur sévérité. Un travail équivalent est en cours pour proposer des lois de transfert locales prenant en considération le nombre de cycles et leurs caractéristiques. Il s’agira là d’une avancée considérable par rapport aux méthodes forfaitaires disponibles. Une première présentation synthétique des acquis du projet SOLCYP sera publiée dans les actes de l’atelier du TC 209 de l’ISSMGE à l’occasion de la 18e ICSMGE. Un ouvrage de recommandations professionnelles sur le calcul des pieux sous sollicitations cycliques sera publié à l’issue du projet.

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CONCLUSION

L’innovation que représentent les Projets Nationaux de recherche en France ne résulte pas seulement du fait que la majorité du financement est fournie par les partenaires. En effet pour qu’un projet puisse être réalisé, il est aussi nécessaire qu’il réunisse un nombre suffisant de partenaires et pour cela que son thème de recherche réponde à l’attente de l’ensemble de la profession dans la branche correspondante du génie civil, par exemple la géotechnique. C’est grâce à l’IREX, l’organisme de gestion des Projets Nationaux, et à ses pôles de compétence que de tels thèmes fédérateurs sont trouvés et proposés. En outre les moyens expé